Chapter 22 - Water and Waste Water Treatment Plant Hydraulics

Source: HYDRAULIC DESIGN HANDBOOK

CHAPTER 22

WATER AND WASTEWATER TREATMENT PLANT HYDRAULICS Federico E. Maisch Sharon L. Cole David V. Hobbs Frank J. Tantone William L. Judy Greeley and Hansen Richmond, VA

22.1 INTRODUCTION Designers of water treatment plants and wastewater treatment plants are faced with the need to design treatment processes which must meet the following general hydraulic requirements: • Water treatment plants. Provide the head required to allow the water to flow through the treatment processes and to be delivered to the transmission/distribution system in the flow rates and at the pressures required for delivery to the users. • W Wastewater treatment r plants. Provide the head required to raise the flow of wastewater from the sewer system to a level which allows the flow to proceed through the treatment processes and be delivered to the receiving body of water. The above requires knowledge of open-channel, closed-conduit, and hydraulic machine flow principles. It also requires an understanding of the interaction between these elements and their impact on the overall plant (site) hydraulics. Head is either available through the difference in elevation (gravity) or it has to be converted from mechanical energy using hydraulic machinery. Distribution of flows using open channels or closed conduit is critical for proper hydraulic loading and process performance. This chapter brings together information on commonly used hydraulic elements and specific applications to water treatment plants and wastewater treatment plants. The development of hydraulic profiles through the entire treatment process with examples for water treatment plants and wastewater treatment is also presented. Many processes and flow control devices are similar in both water treatment plants and wastewater treatment plants. Both types of plants require flow distribution devices, gates and valves, and flowmeters. These devices are discussed in Section 22.2. The development of water treatment plant hydraulics, including examples from in-place facilities, are presented in Section 22.3. Wastewater treatment plant hydraulics are discussed in Section 22.4, and Section. 22.5 is devoted to non-Newtonian flow principles. 22.1 Downloaded from Digital Engineering Library @ McGraw-Hill (www.digitalengineeringlibrary.com) Copyright © 2004 The McGraw-Hill Companies. All rights reserved. Any use is subject to the Terms of Use as given at the website.

WATER AND WASTEWATER TREATMENT PLANT HYDRAULICS

22.2

Chapter Twenty-Two

22.2 GENERAL 22.2.1 Introduction This section addresses some elements which are common to both water treatment plants and wastewater treatment plants including: • Flow distribution–manifolds • Gates and valves • Flowmeters • Local losses

22.2.2 Flow distribution–manifolds In the design of water and wastewater treatment plants, proper flow distribution can be as critical as process design considerations, which typically receive much more attention. Plant failures resulting from unequal and unmanageable flow distribution are possibly as common and as serious as those resulting from errors in process design. Flow distribution devices, such as distribution channels, pipe manifolds or distribution boxes, are commonly used to distribute or equalize flow to parallel treatment units, such as flocculation tanks, sedimentation basins, aeration tanks, or filters. 22.2.2.1 Distribution boxes. The simplest of these devices, the distribution box, typically consists of a structure arranged to provide a common water surface as the supply to two or more outlets. The outlets are typically over weirs and the key to equal flow distribution is to provide independent hydraulic characteristics between the downstream system and the water level in the distribution box. In other words, provide a free discharge weir (nonsubmerged under all conditions) for each outlet to eliminate the impact of downstream physical system differences on the flow distribution. Velocity gradients across the distribution box must be nearly zero to equalize flow conditions over each outfall weir. Weirs clearly should be of uniform design in terms of physical arrangement length and materials of construction. They should also be adjustable to account for any minor flow differences noted in actual operation. The same principles apply if the designer wishes to distribute flows in specific proportions which are not necessarily equal. In this case the designer could control the proportions of flow distribution by varying the relative geometry of the weirs (i.e., change the width or invert of each weir to achieve a desired flow distibution). The specifics of weir hydraulics are covered in various texts in the literature. Attention should always be paid to the selection of the proper coefficients to model the specific weir geometry and the geometry of the approach flow. 22.2.2.2 Distribution channels and pipe manifolds. Distribution channels and manifolds are also common in plant design but a bit more complex in their function and design. The distribution of flow in these devices is impacted by the flow distribution itself. Since a portion of the flow leaves the channel or manifold along the length of the device, the velocity of flow and, therefore, the relationship of energy grade line, velocity head and hydraulic grade line varies along the length of the device. This is more clearly visible in a distribution channel of uniform cross section, using side weirs along its length for flow distribution. At each weir, flow leaves the channel, resulting in less velocity head in the channel

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Water and Wastewater Treatment Plant Hydraulics 22.3

and possibly a higher water surface at each ensuing weir. Chao and Trussell (1980), Camp and Graber (1968), and Yao (1972) have presented comprehensive approaches for the design of distribution channels and manifolds and should be reviewed for details of design. As in distribution boxes, the most important consideration to achieving equalized flow distribution is to minimize the effects of unequal hydraulic conditions relative to each point of distribution. In channels this can be accomplished by tapering the channel cross section, varying weir elevations, making the channel large enough to cause velocity head changes to be insignificant or a combination of these. Similar considerations may be applied to manifolds with submerged orifice outlets. A reliable approach here is to provide a large enough manifold, resulting in a total headloss along the length of the distribution of less than one tenth the loss through any individual orifice. This approach essentially results in the orifices becoming the only hydraulic control and the accuracy of the flow distribution is then dependent on the uniformity of the orifices themselves.

22.2.3 Gates and Valves Gates and valves generally serve to either control the rate of flow or to start/stop flow. Gates and valves in treatment plants are typically subjected to much lower pressures than those in water distribution systems or sewage force mains and can be of lighter construction. 22.2.3.1 Gates. Gates are typically used in channels or in structures to start and stop flow or to provide a hydraulic control point which is seldom adjusted. Because of the time and effort required to operate gates, they are not suited for controlling flow when rapid response, frequent variation, or delicate adjustments are needed. Primary design considerations when using gates are the type of gate fabrication and the installation conditions during construction. There are many fabrication details including materials used, bottom arrangement, and stem arrangement. For instance, for solids bearing flows, a flush bottom, rising stem gate can be used to avoid creating a point of solids deposition and to minimize solids contact with the threaded stem. Gate manufacturers are a good source of information for gate fabrication details and can assist with advice regarding specific applications. Most commonly used gates are designed to stop flow in a single direction. They may use upstream water pressure to assist in achieving a seal (seating head), but typically also must be designed to resist static water pressure from downstream (unseating head). Both seating and unseating heads must be evaluated in design of a gate application. For most manufacturers, the seating or unseating head is expressed as the pressure relative to the center line of the gate. 22.2.3.2 Valves. Table 22.1 provides a summary of several types of valves and their applications. Valves are used to either throttle (control) flow or start/stop flow. Start/stop valves are intended to be fully open or fully closed and nonthrottling. They should present minimum resistance to flow when fully open and should be intended for infrequent operation. Gate valves, plug valves, cone valves, ball valves, and butterfly valves are the most common start/stop valve selections. Butterfly valves have a center stem, are most common in clean water applications and should not be used in applications including materials that could hang-up on the stem. Therefore, they are seldom used at wastewater plants prior to achieving a filter effluent water quality.



22.4

Chapter Twenty-Two

TABLE 22.1

Typical Valves and Their Application*

Type

Open/Close

Throttling

Water

Wastewater

Sluice gate

X

X

X

Slide gate

X

X

X

Gate valve

X

X

X

Plug valve

X

X

X

X

Cone valve

X

X

X

X

Ball valve

X

X

X

X

Butterfly valve

X

X

X

Swing check

X

X

Lift check

X

X

X

Ball check

X

X

X

Spring check

X

X

X

Globe valve

X

X

Needle valve

X

X

Angle valve

X

X

X

X

Pinch/diaphragm

X

X

*

Typical applications–exceptions are possible, but consultation with valve manufacturers is recommended.

Check valves are a special case of a start/stop valve application. Check valves offer quick, automatic reaction to flow changes and are intended to stop flow direction reversal. Typical configurations include swing check, lift check, ball check and spring loaded. These valves are typically used on pump discharge piping and are opened by the pressure of the flowing liquid and close automatically if pressure drops and flow attempts to reverse direction. The rapid closure of these valves can result in unacceptable “waterhammer” pressures with the potential to damage the system. A detailed surge analysis may be required for many check valve applications (see Chapter. 12). At times, mechanically operating check valves should be avoided in favor of electrically or pneumatically operated valves (typically plug, ball, or cone valves) to provide a mechanism to control time of closing and reduce surge pressure peaks. Throttling valves are used to control rate of flow and are designed for frequent or nearly continuous operation depending on whether they are manually operated or electronically controlled. Typical throttling valve types include globe valves, needle valves, and angle valves in smaller sizes, and ball, plug, cone, butterfly, and pinch/diaphragm valves in larger sizes. Throttling valves are typically most effective in the mid-range of loose line open/close travel and for best flow control should not be routinely operated nearly fully closed or nearly fully open.

22.2.4 Flow meters The most common types of flow meters used in water and wastewater treatment plants are summarized in Table 22.2 and fall into the following categories:



Water and Wastewater Treatment Plant Hydraulics 22.5 TABLE 22.2 Type

Common Types of Flow Meters Typical Accuracy

Size Range

Headloss

Cost

W

WW

Venturi

0.75% of rate

1–120 in

Low

Medium

X

X

Orifice plate

2% of scale

Any size

Medium

Low

X

X

Pitot tube

0.5–5% of scale

1/2–96 in

Low

Low

X

Parshall flume

5% of rate

Wide range

Low

Medium

X

X

Magnetic

0.5% of rate

1/10–120 in

None

High

X

X

Doppler

1–2.5% of rate

1/8–120 in

None

High

X

X

Propeller

2% of rate

Up to 24 in

High

High

X

Turbine

0.5–2% of rate

Up to 24 in

High

High

X

• Pressure differential/pressure measuring meters (e.g., Venturi, orifice plate, pitot tube, and Parshall flume meters) • Magnetic meters • Doppler (ultrasonic) meters • Mechanical meters (e.g., propeller and turbine meters) Accurate flow measurements require uniform flow patterns. Most meters are significantly impacted by adjacent piping configurations. Typically a specific number of straight pipe diameters is required both upstream and downstream of a meter to obtain reliable measurements. In some cases, 15 straight pipe diameters upstream and 5 straight pipe diameters downstream are recommended. However, different types of meters have varying levels of susceptibility to the uniformity of the flow pattern. Meter manufacturers should be consulted. 22.2.4.1 Pressure differential/pressure measuring meters. Pressure differential/pressure measuring flow meters include Venturi meters, orifice plates, averaging pitot meters, and Parshall flumes. These meters measure the change in pressure through a known flow cross section–or in the case of the pitot meter, measure the difference in pressure at a point in the flow versus static pressure just downstream in a uniform section of pipe. Venturi meters and orifice plates are commonly used in water and wastewater. Solids in wastewater could plug the openings of a pitot tube meter-limiting their use to relatively clean liquids. The Venturi meter and orifice plate meter use pressure taps at the wall of the device and can be arranged to minimize potential for debris from clogging the taps. The Parshall flume can be arranged with a side stilling well and level measuring float system or an ultrasonic level sensing device to measure water level. 22.2.4.2 Magnetic meters. In a magnetic flowmeter, a magnetic field is generated around a section of pipe. Water passing through the field induces a small electric current proportional to the velocity of flow. Because a magnetic meter imposes no obstruction to the flow, it is well suited to measuring solids bearing liquids as well as clean liquids and produces no headloss in addition to the normal pipe loss. Magnetic meters are among the least susceptible to the uniformity of the stream lines in the approaching flow.



22.6

Chapter Twenty-Two

22.2.4.3 Ultrasonic meters. In an ultrasonic flow meter, a pair of transceivers (transmitter/receiver) are positioned diagonally across from each other on the pipe wall. The transmitter sends out a signal which is affected by the speed of the flow. The receiver measures the difference between the speed of the signal when directed counter to the flow (slowed by the flow) and when directed with the flow (speeded up by the flow). The time difference is a function of fluid velocity, which is used to compute the flow. As with magnetic meters, no flow obstruction is imposed resulting in no headloss in addition to the normal pipe loss. Ultrasonic meters are also available for open-channel applications. 22.2.4.4 Mechanical meters. Mechanical meters include propeller and turbine-type equipment. The two meters are similar in function in that in each a device is inserted into the flowpath. The device is rotated by the flow and the speed of rotation is used to compute rate of flow. These devices impose an obstruction to flow, are recommended for clean water only, and generally result in significant headloss.

22.2.5 Local Losses In any piping system as flow travels along the pipe, pressure drops as a result of headloss due to friction along the pipe and local losses at bends, fittings, and valves. The local losses at bends, fittings, and valves are least significant in long, straight piping systems and most significant at treatment plants where the length of straight pipe is relatively short and therefore, the frictional pipe losses comprise a smaller fraction of the total losses when compared to the summation of all local losses. A term often used to refer to local losses is “minor losses,” however, because of the later consideration the term “minor losses” can be misleading. Traditionally, local losses have been computed in terms of “equivalent length” of straight pipe or in terms of multiples of velocity head. The “equivalent length” or loss factor K methods attempt to estimate the local losses based on the characteristic of the specific bend, fitting or valve. The K loss factor method is discussed here. Essentially, a local loss is computed as follows: 2 hL KV 2g

(22.1)

where hL local loss, K loss factor, V velocity, g gravitational acceleration. The values for K reported by various sources vary considerably for some local losses and are relatively consistent for others. See references. There are many literature sources for K values. The Bureau of Reclamation (1992) is one such source of information regarding energy loss equations. Table 22.3 shows a range of K factors from additional sources as well as a typically used value for each. Judgment must be applied in computing local losses, taking into account any unique system conditions. Throughout this chapter K values were obtained from equipment manufacturers when available. Values from Table 22.3 were used only as an approximation when more specific data were unavailable. The reader is cautioned that there are application-specific characteristics which have significant influence on the K factors. One of these characteristics, for example, is size. A K value of 0.6 is often encountered in literature to characterize the losses associated with flow through the run of a tee. However, for flow past tees in large pipes this factor can be very small and nearly zero.




Gate valve 100% open 0.39 75% open 1.1 50% open 4.8 25% open 27 Globe valve–open 10 Angle valve–open 4.3 Check valve–ball 4.5 Swing check Butterfly valve–open 1.2 Foot valve–hinged 2.2 Foot valve–poppet 12.5 Elbows 45° regular 45° long radius 90° regular 90° long radius 180° regular 180° long radius (flanged) Tees Std. teee–flowthrough run 0.6 Std. teee–flow-through branch1.8 Return bend 1.5 Mitre bend 90° 1.8 60° 0.75 30° 0.25 Expansion d/D = 0.75 0.18 d/D = 0.5 0.55 d/D = 0.25 0.88 Contraction d/D = 0.75 0.18 d/D = 0.5 0.33 d/D = 0.25 0.43 Entrancee–projecting 0.78 Entrancee–sharp 0.5 Entrancee–well rounded 0.04 Exit 1.0

0.19 1.15 5.6 24 10 5

2.1–3.1 65–70 0.6–2.3

0.30–0.42 0.18–0.20 0.21–0.3 0.14–0.23 0.38 0.25

0.19

0.1–0.3

10 5

4.0–6.0 1.8–2.9 06–2.2 0.16–0.35 1.0–1.4 5.0–14.0

0.6–2.5

0.18 0.25 0.18

0.5 0.7 0.6

0.6 1.8 1.8 2.2

0.3 0.75 0.4

1.8

1.129–1.265 0.471–0.684 0.130–0.165

0.8 0.35 0.1

Typically Used Value

Committee on Pipeline Planning (1975) 0.2 1.2 5.6 24 10 2.5

0.42

0.6 1.8 2.2

0.78 0.5 0.04 1.0

Sanks (1989)

Simon (1986)

Cameron Hydraulic Data

Daugherty (1977)

Crane Co. (1987)

Walski (1992)

Valve and Fitting Types

Bulletin No. 2552, University of Wisconsin

Typical K Factors for Computing Local Losses

Ten-State Standards (1978)

TABLE 22.3

0.2 1.2 5.6 25 10 5 5 2.5 0.5 2.2 14 0.42 0.2 0.25 0.19 0.38 0.25 0.6 1.8 2.2 1.3 0.6 0.16

0.19 0.56 0.92

0.2 0.6 0.9

0.2 0.6 0.9

0.19 0.33 0.42 0.83 0.5

0.2 0.3 0.4 0.78 0.5 0.04 1.0

0.2 0.33 0.43 0.8 0.5 0.04 1.0

0.8 0.5 0.04 0.04

0.8 0.5 0.25 1.0



22.8

Chapter Twenty-Two

22.3 HYDRAULICS OF WATER TREATMENT PLANTS 22.3.1 Introduction Water treatment comprises the withdrawal of water from a source of supply and the treatment of raw water through a series of unit processes for the beneficial use of the system customers. Raw water quality can vary widely. The ultimate uses of water by the system customer (e.g., drinking, fire protection, irrigation, aquifer recharge, etc.) can also vary and be subject to different treatment level requirements and regulations. Therefore, the selected treatment processes vary widely over a multitude of treatment technologies in use. Water treatment consists of a series of chemical, biological, and physical processes connected by channels and pipelines. Figures 22.1 and 22.2 illustrate process flow diagrams (flowsheets) for typical surface water and groundwater treatment plants, respectively. The designer of the water treatment process must carefully evaluate source water characteristics and desired water quality characteristics of the treated water to design treatment processes capable of purifying the source water to water suitable for the system customers. The objective of this chapter is to review the hydraulic considerations required to convey water through the treatment process. Design of a plant’s treatment process is closely linked with the hydraulic design of the treatment plant. This chapter presumes that the designer has evaluated and selected treatment processes for the water treatment plant. Although design flows are discussed below, we have also assumed that the designer has chosen a design flow requirement for the treatment process. For municipal treatment plants, design flows are based on the service area

FIGURE 22.1 Typical surface water treatment plant process flow diagram.

FIGURE 22.2 Typical ground water treatment plant process flow diagram with dual trains (#1 and #2).




population and the per capita use of water by the population served. The per capita use of water can be obtained from literature sources as an initial approximation. However, these initial estimations must be corroborated with actual site specific population counts and water usage. For nonmunicipal treatment facilities, treated water needs of the service area must be individually evaluated. 22.3.1.1 Sources of supply. Natural sources of supply include groundwater and surface water supplies. Groundwater supplies typically are smaller in daily delivery but serve more systems than surface water supplies. Groundwater supplies normally come from wells, springs, or infiltration galleries. Wells constitute the largest source of groundwater. Except in rare circumstances of artesian wells (wells under the influence of a confined aquifer) and springs, groundwater collection involves pumping facilities. Hydraulics of groundwater treatment plants are frequently based on hydraulics of conduits under pressure, such as pipelines, pressure filters, and pressure tanks. Raw water characteristics of groundwaters are uniform in quality compared with surface supplies. Surface water supplies are normally larger in daily delivery. Surface supplies are used to service larger population centers and industrial centers. In areas where groundwater supplies are limited in yield or where groundwater supplies contain undesirable chemical characteristics, smaller surface water treatment plants may be utilized. Surface water sources of supply include rivers, lakes, impoundments, streams, and ponds. The treatment processes chosen in plants treating surface water favor nonpressurized systems such as gravity sedimentation. The larger flow volumes characteristic of surface water supplies also favor open channel hydraulic structures for conveying water through the treatment process. Raw water characteristics of surface supplies can vary rapidly over short periods of time and also experience seasonal variation. 22.3.1.2 Treatment requirements. Treatment requirements for municipal water treatment plants are normally defined by regulatory agencies having authority over the plant’s service area. In the United States, regulatory agencies include national government regulations promulgated through the Environmental Protection Agency and state government regulations. Water treatment plants are designed to meet these regulations. Treatment regulations change through improved knowledge of health effects of water constituents and through identification of possible new water-borne threats. The designer therefore should attempt to select treatment processes which will also meet treatment requirements which are expected to be promulgated over the next few years. To the extent possible, treatment plant process design should provide flexibility for future plant expansions or for possible additional treatment processes. Because hydraulic design of plants must go hand-in-hand with the process selection, plant hydraulic design should provide for the flexibility to add future plant facilities. Treatment requirements for industrial water treatment plants are dictated by process needs of the industry and less by regulatory agency requirements. Industrial water treatment plants that result in contact between or ingestion of the treated water by humans must conform to the local regulatory requirements. 22.3.1.3 General design philosophy. Effective design of water treatment plant hydraulics requires that the hydraulic designer have a thorough knowledge of all aspects of the water system. The overall treatment system hydraulic design must be integrated and coordinated including the treatment plant, the raw water intake and pumping facilities, the treated water storage, and treated water pressure/head requirements. The design within the water treatment plant must also be integrated between the various treatment processes.



22.10

Chapter Twenty-Two

Additionally, design considerations must address the availability of operating personnel and hours of operation such that the process and hydraulic requirements conform to available resources.

22.3.2 Hydraulic Design Considerations in Process Selection Water treatment plant process selections are controlled principally by characteristics of the raw water and by the desired water quality characteristics of the finished water. Flow through each unit process and each conduit connecting processes results in loss of hydraulic head. Most treatment plants have limited head available. The selection of a particular unit process will include evaluation of numerous criteria including costs, operability, performance, energy use and similar items. One criteria which must also be evaluated for each process is the hydraulic head requirements of the process. 22.3.2.1 Head available. For the design flow to pass through a water treatment plant, the total available head must exceed the head requirements of the unit processes and connecting conduits. The head available is the difference in energy grade line (EGL) in the hydraulic profile between the head works of the plant and the end of the plant. Additional head may be provided by pumping or by lowering the elevation of treatment units at the end of the plant. See Figure 22.3 for a typical water treatment plant hydraulic profile. For most surface water plants, the hydraulic profile at the head of the plant is controlled by raw water pumps pumping from the intake facilities. The hydraulic profile at the head of a plant in a groundwater system is typically determined by the well pumps serving the plant. 22.3.2.2 Typical unit process head requirements. Following below is a table of typical head requirements for water treatment plant processes. This table may be used for initial evaluation of unit processes. More detailed hydraulic evaluations must be performed after plant operating modes and design flows are determined. Detailed hydraulic evaluations must also include headlosses in connecting conduits.

Unit Process Intakes, including bar screens Rapid mixing Flocculation Sedimentation Filtration – Gravity – Pressure Disinfection Aeration – Spray – Cascade – Compressed air Softening Ion exchange softening Iron and manganese removal

Head Requirement at Rated Capacity, m (ft) 0.3–0.9 (1–3) 0.15–0.30 (0.5–1) 0.06–0.15 (0.2–0.5) 0.6–2.4 (2–8) 3–4.6 (10–15) 3–7.6 (10–25) 0.15–0.6 (0.5–2) 3–4.6 (10–15) 3–4.6 (10–15) 0.15–0.6 (0.5–2) 0.15–0.6 (0.5–2) 0.6–1.5 (2–5) 0.6–1.5 (2–5)



FIGURE 22.3 Hydraulic profile.




22.12

Chapter Twenty-Two

22.3.3 Hydraulic Design Considerations in Plant Siting Plant sites are normally selected before the hydraulic designer initiates design of the treatment system. If a plant site has not been selected, the designer should be aware of hydraulic considerations which may influence site selection. Site elevation has the most significant impact on plant hydraulics. A plant site located above the service area will eliminate or reduce pumping requirements from the plant to the service areas. Typical municipal distribution system pressures are 40–70 psi, therefore the elevation of the treatment plant should be at least 100 ft above the service area to eliminate finished water pumping. Similarly, plant sites which permit gravity intake of the source water may reduce or eliminate raw water pumping. Few plants are able to meet these optimal conditions. The typical surface water plant must pump both raw and finished water. Raw water (low-lift) pumps are used to pump water from the water source into the treatment facilities and finished water (high-lift) pumps are used to pump from the treatment plant into the service area distribution system.

22.3.4 Hydraulic Design Consideration in Plant Layout After the plant site has been identified, the plant design may be arranged for optimal hydraulic benefit. In particular, arrangement of treatment processes to allow flow to move down gradient minimizes excavation needs for structures. Arrangements which are designed for future expansion should consider the hydraulic needs of the expanded plant as well as the process needs. Grouping of processes together facilitates movement of water through the treatment process train. The designer should also consider secondary hydraulic systems for optimal design. Chemical feed systems and dewatering systems are examples of secondary hydraulic systems which must be coordinated with the treatment flow system. Normally it is desirable to minimize the length of chemical piping systems. Dewatering systems are usually based on gravity drainage of basins and conduits.

22.3.5 Bases for Design After evaluation and selection of a source of supply and development of the treatment plant process train, the designer is prepared to develop the plant Bases for Design. The Bases for Design is a summary of design flow and capacity, and proposed treatment processes, including the chemical storage and feed facilities. 22.3.5.1 Design flows. Design flows for water treatment plants serving municipalities are typically based on the projected population within the water service area for the design life of the treatment facilities. Population data is normally determined from census records, land use zoning information, and studies of existing and projected population densities. Service area per capita demands are affected by the mix of domestic, commercial, and industrial water users which are unique to each service area.Typically water consumption records are available for water service areas. For new facilities, the use of generalized water consumption data may be needed. In the United States, water consumption varies widely but generally ranges between 100–200 gallons per capita per day.




From studies of projected population and per capita demand, planned design flows for the water treatment facilities may be developed. These demands include the following: • Annual average demand. The average daily water consumption for the water service areas, generally computed by multiplying the average daily consumption (gallons per capita) by the projected population of the service area. • Maximum demand. Maximum demand experienced by the water plant throughout its service life. The maximum hour demand is generally 200 to 300 percent of the average demand but numerous factors affect the peak demand experienced by water treatment plants. These factors include seasonal demands (particularly for plants where service areas are located in extremes of hot and cold temperatures), normal daily flow variations, the community size, industrial usage, and system storage. Normally system storage is provided to service peak hour demands, allowing the treatment facilities to be designed on peak day demands. Peak day demands generally range between 125 and 200 percent of the average demand. • Minimum flow. As the name suggests, the minimum flow expected to be processed through the treatment facilities. Minimum flow depends upon system operations. In general, minimum flows for municipal plants may be estimated as 50 percent of the average demand, but range between 25 and 75 percent of the average demand. 22.3.5.2 Rated treatment capacity. The rated treatment capacity of a plant is that capacity for which each of the unit processes are designed. For municipal treatment plants with adequate system storage, the rated treatment capacity is the system’s maximum day demand. Where storage is limited, the rated treatment capacity may be greater, for example, the system maximum hour demand or greater. Smaller systems may be designed to produce the rated treatment capacity in one or two 8-h shifts rather than over the entire 24-h day. 22.3.5.3 Hydraulic treatment capacity. Treatment plants are normally designed for a hydraulic capacity greater than the rated treatment capacity. Hydraulic treatment capacities are normally equal to 125 to 150 percent of the rated treatment capacity. The hydraulic treatment capacity provides flexibility for future process changes or alternative flow routings through the plant. Hydraulic capacities in excess of the rated treatment capacity provide some margin of safety for operations which may not be optimal (e.g., control gates inadvertently left partially open). 22.3.5.4 Treatment process bases for design. The development of the water treatment plant’s “Bases for Design” is a key step in establishing the criteria to which the plant will be designed. This document must be reviewed carefully with the water treatment plant owner representatives and understood and agreed to by all before the final design proceeds. The Bases for Design presents a summary of each treatment process including design flows (minimum, average, rated capacity), specification of dimension of major elements (e.g., tanks, pumps), both hydraulic and process loading characteristics, required performance, and design data for the chemical storage and feed system. Table 22.4 presents an example of the bases for design for sedimentation basins (one of the many unit processes in a water treatment plant).

22.3.6 Plant Hydraulic Design As noted above, a water treatment plant consists of a series of treatment processes connected by free surface flow channels and pipelines. During development of the plant’s



22.14

Chapter Twenty-Two

TABLE 22.4

Treatment Process Bases for Design—Sedimentation Basins Item

Stage I

Stage II

StageIII

Annual Average

Maxi– mum Day

Annual Average

Maxi– mum Day

Annual Average

Maximum Day

4

4

8

8

12

12

3.17

1.99

3.17

1.99

3.17

1.99

Surface loading [(galⴢm)/ft ]

0.47

0.75

0.47

0.75

0.47

0.75

Flowthrough velocity (ft/min)

1.21

1.93

1.21

1.93

1.21

1.93

8

8

8

8

8

8

1

1

1

1

1

1

4

4

4

4

4

4

Number of basins Basin characteristics Plan–75 ft 230–6 in Nominal side water depth–12 ft (SWD) Surface area/basin–17,288 ft2 Volume/basin–207,456 f3 Channels/basin–2 L:W ratio–6.1:1 Displacement time (h) 2

Sludge collectors Longitudinal collectors Type: chain flight Number per basin Cross collectors Type: chain flight Number per basin Settled sludge pumps Type: progressive cavity Number: 100 gal/min capacity 400 gal/min capacity

4

4

4

4

4

4

200 gal/min capacity

—

—

8

8

16

16

Capacity (gal/min) Installed

2000

2000

3600

3600

5200

5200

Firm

1600

1600

3200

3200

4800

4800

Bases for Design, the designer determines the rated treatment capacity, average flow, minimum flow and hydraulic capacity of the plant. Following development of the Bases for Design, the designer must evaluate plant operating modes to develop a detailed plant flow diagram and hydraulic profile through the plant.




22.3.6.1 Plant operating modes. Operating modes describe the sequence of treatment processes the water goes through to achieve the required level of purification. Operational modes are normally presented in the form of simplified block diagrams which illustrate the flow path through the plant from one process to the next. These operational mode block diagrams are useful in visualizing stages during construction, future planned plant expansions or simply alternative operating modes. Figures 22.4 through 22.9 show an example of a sequence of plant operating modes for a surface water treatment plant which illustrate three stages of a plant expansion program with alternatives for the flocculation and sedimentation basins to work in series or in parallel. Plant processes proposed include raw water control chambers, rapid mix chambers, flocculation/sedimentation basins, ozone contact chambers, and filters. In this example, the raw water control chambers are used to split flow between plant process groups and also as a rapid mix chamber for chemical addition.

FIGURE 22.4 Stage I—operational mode diagram.

FIGURE 22.5 Stage II—parallel operational mode diagram.



22.16

Chapter Twenty-Two

The Stage I facilities including raw water control chamber, flocculation/sedimentation basins and filters are depicted in Fig. 22.4. Operational modes for a proposed plant expansion to double the plant capacity (Stage II) are shown in Figs. 22.5 through 22.7 and operating modes for a second plant expansion to triple the plant capacity (Stage III) are shown in Figs. 22.8 and 22.9. Settled water ozone contact chambers were added to the expanded plant, which illustrates treatment upgrades. Operational modes for the Stage II treatment plant include parallel and series flocculation/sedimentation. When the plant is operated in the parallel mode, influent raw water for each set of sedimentation basins flows by gravity from the raw water control chamber serving the basin set. Raw water flow is divided between each sedimentation basin in service at the raw water control chamber. Settled water from each set of basins is routed to an ozone contact chamber. Ozonated settled water is then combined prior to flowing to the filters.

FIGURE 22.6 Stage II—series flocculation/sedimentation basin operational mode diagram.

FIGURE 22.7 Stage II—split parallel operational mode diagram.




FIGURE 22.8 Stage III—parallel operational mode diagram.

FIGURE 22.9 Stage III—split parallel operational mode diagram.

The Stage III split parallel operational mode is similar to the parallel operational mode except that the ozonated settled water from each set of basins is not combined prior to flowing to the filters. Side-by-side plant scale treatment studies are possible with the future split parallel mode since part of the flocculation/ sedimentation/filtration processes can be operated as a “control” while the remainder of the plant can be operated in a controlled experimental mode. The series flocculation/sedimentation operational mode is designed to permit operation of the sedimentation basins in two stages in lieu of the single–stage parallel mode. Under certain raw water conditions, operation of the basins in series may enhance performance of the basins. Chemical feed for the first and second sedimentation stages may be adjusted to respond to raw water conditions and settled water quality after the first–stage sedimentation. Series flocculation/sedimentation increases hydraulic losses through the



22.18

Chapter Twenty-Two

plant. Under this mode, twice as much flow is routed to each basin and the flow pattern is longer, since the settled water from the first sedimentation stage must be returned to the influent of the second sedimentation stage. Operational mode block diagrams are also a convenient means to illustrate the effect of side stream flows which may impact the overall plant flow. For example, removal of sludge from the sedimentation basins is accompanied by a decrease in flow leaving the basins compared with flow entering the basins. In a similar manner, filter backwash water removes a certain amount of flow. A plant designed to produce a certain rated capacity may have to treat more than the rated capacity through certain processes. The impact of these side stream flows must be evaluated on an individual basis. In many treatment plants, backwash water treatment facilities are installed to recycle backwash water to the head of the plant. 22.3.6.2 Plant flow diagrams. After establishing plant operating modes, more detailed flow diagrams are developed by the designer. The diagrams normally start with possible valving and gating arrangements and are then expanded with tentative valve, sluice gate, pipeline, and conduit sizes. Valving arrangements are designed to enable any of the major operational units (e.g., sedimentation basin, ozone contact chamber) to be removed from service. The arrangement may include design of temporary flow stop devices, such as stop logs (sectional barriers which were originally constructed of logs but are now commonly metal plates). The arrangement should be designed to permit maintenance work on major valves and sluice gates while minimizing the impact on plant process. Major channel sections should be designed so they can be removed from service and dewatered while minimizing impacts on the rest of the plant. The designer should distinguish between units taken out of service frequently (such as filters), periodically (such as sedimentation basins), or rarely (such as conduits). Filter backwashing occurs so frequently that the rated treatment capacity can be met with one filter out for backwashing. Sedimentation basins may be removed from service once or twice per year for equipment maintenance. Since the basins outages occur at widely scattered intervals, it is reasonable to design the units to be removed from service during lower flow periods. Conduits and pipelines are rarely removed from service, but the hydraulic impacts can be significant. Depending on the conduit location, removal of a conduit can remove a portion of the plant from service. Effective design will provide redundant conduits so that a portion of the plant can remain in service during conduit dewatering. The focus of this section has been on the main plant hydraulics, but the hydraulic designer must also design for hydraulic subsystems. An important group of these subsystems include dewatering of all basins and conduits. Where plant elevations will allow, gravity dewatering is recommended. In most cases, dewatering pumps are necessary. These pumps may be located in the unit being dewatered or may be located in a separate structure connected to the process unit by dewatering pipelines. 22.3.6.3 Hydraulic Profile. One of the most important tools in the hydraulic design of a water treatment plant is the development of a hydraulic profile. The hydraulic profile is a diagram showing the energy grade line (EGL) at each unit process. For open tanks with flows at minimal velocities, which is the case in most water treatment plants, the velocity head is negligible and the hydraulic grade line (HGL) or water surface elevation (WSEL) provide an adequate representation of the EGL. Profiles normally include critical structural elevations of processes and conduits. The profile may also include ground surface profiles and other site information.




Hydraulic profiles are developed for each of the design flows. In the case of water treatment plants, the design flows may include rated treatment capacity, hydraulic capacity, average flow, and minimum flow. Hydraulic profiles should also take into consideration unit processes or conduits which may be taken out of service. Hydraulic profiles are valuable design and operational tools to assist in scheduling routine maintenance activities and for evaluating the impact to the treatment plant capacity during outages of process units or conduits. Computations of hydraulic profiles begin at control points where there is a definite relationship between the plant flow and water surface depth. For gravity flow plants, the most common forms of control points are weirs and tank water surface elevations (e.g., clear well water surface elevations), but other types of control points may be used. From each control point, head losses associated with local losses, plant piping, and open channel flow are added to the control water surface. Since flows in water treatment plant’s are mostly in the subcritical regime (Froude number 1), most hydraulic designers will work upstream from the control point. For pressure plants, control points are typically pressure regulating or pressure control points, frequently in the service area distribution system. From these control points and knowledge of the flow velocity, both the EGL and HGL may be computed back to the treatment facilities. Hydraulic profiles are valuable design tools to identify overall losses through the plant. Profiles are also valuable to identify units with excessive losses. Since total head available is normally limited, units with excessive losses should be considered for redesign to reduce local loss coefficients or to reduce velocities. Figure 22.3 is an example hydraulic profile for a gravity surface water treatment plant with conventional treatment processes. The method of computing headlosses is presented in Section 22.3.7.

22.3.7 Water Treatment Plant Process Hydraulics In this section calculations required to establish the WSEL through a medium-sized water treatment plant will be presented. A schematic of the water treatment plant is shown in Fig. 22.10. Notice that future growth has been considered in the initial design. Three examples are included which illustrate typical hydraulic calculations. The first example calculates the WSEL from the sedimentation basin effluent chamber back through the flocculation/sedimentation basins to the Raw Water Control Chamber. The second follows the flow from the clear well back through the filters. Filter hydraulics are illustrated in the third example. All examples are presented in a spreadsheet format which is designed to facilitate calculating the EGL, HGL, and WSEL at various points through the treatment process and for multiple flow rates (i.e., minimum, daily average, peak hour, future conditions). 22.3.7.1 Coagulation. Process criteria and key hydraulic design parameters. The coagulation process, used to reduce particulates and turbidity, is carried out in three steps: mixing (often referred to as rapid or flash mixing), flocculation, and sedimentation. Each of these steps is briefly discussed below. Rapid mixing. The mixing process imparts energy to increase contact between existing solids and added coagulants. Possible mixer types include turbine, propeller, pneumatic, and hydraulic. Headloss that occurs in mixing chambers depends on the chosen mixing device. Most mechanical mixers do not create significant head losses. The headloss coefficient (K) K associated with a specific mixer can be obtained from the manufacturer. Pneumatic mixing, which is not common, has associated losses similar to those



22.20

Chapter Twenty-Two

FIGURE 22.10 Schematiz of a water treatment plant.

for aeration (see table in Section 22.3.2.2, above). Hydraulic mixing takes place using weirs, swirl chambers, throttled valves, Parshall flumes, or other devices to induce turbulence. Head loss coefficients for these devices can be obtained from the manufacturer. Important considerations during the initial design of a mixing chamber include: • Velocity gradient. This is mixer—specific information and can be obtained from the manufacturer. The system should be designed to provide a velocity gradient that is optimal for the coagulation process taking place. • Dead spots and short circuiting. An ideal mixing system will have minimal dead spots and short circuiting. These can be avoided with proper sizing and placement of mixers. Flocculation. Coagulated particles form larger particles (flocs) during the gentle mixing of flocculation, where the flow travels slowly through a series of flocculator paddles, baffles, or conduits. Inlets and weirs are designed to provide low turbulence for protection of the flocs. The energy provided to the system by the flocculators (manufacturer-specific) or baffling is decreased as the flow approaches the sedimentation basins. Sedimentation. Gravity sedimentation removes coagulated solids prior to filtration. There are four zones in a clarifier as shown in Fig. 22.11 and listed below: • Inlet zone—where upstream flow conditions transition smoothly to uniform flow settling conditions • Sedimentation zone—where sedimentation takes place • Sludge zone—where solids collect and are removed • Outlet zone—where settling conditions smoothly transition to downstream flow conditions




FIGURE 22.11 Hypothetical zones in a rectangular sedimentation basin.

Each of the zones is designed to minimize turbulence and avoid short circuiting. The velocity in the sedimentation zone is limited to 0.3 m/s (1 ft/s) for average flow. Sludge removal equipment moves slowly so that settling patterns are not disturbed. Because the process is designed for smooth flow and minimal turbulence, very little head loss occurs in sedimentation basins. Ports at the inlet and outlet produce the greatest head losses in this process. Hydraulic design example. Table 22.5 illustrates the calculation of the WSEL, using metric units, through the coagulation process at the medium-sized water treatment plant shown in Fig. 22.10. Figs. 22.12 through 22.14 show plan views and details of the

TABLE 22.5

Hydraulic Calculations of a Typical Coagulation Process, SI Units Initial Operation Parameter

1. Plant Flow (m3/s) Note: For Points 1 through 8, see Fig. 22.12 2. WSEL at Point 1 (Calculation done in Table 22.6) (m) 3. Point 1 to Point 2 Average flow 21Q/32 (m3/s) Flow depth WSEL @ 1 – invert (106.60 m) (m) Flow area 5.13 m width depth (m2) Velocity flow/area (m/s) Hydraulic Radius r A/P / (P w 2d) (m) Conduit loss [(V n)/(rr2/3)]2 L (m) where Manning’s n 0.014 and L 28.96 m WSEL at Point 2 (m)

Min. Day. Avg. Day

Design Operation Avg Day Max. Hour

2.19

3.06

3.28

4.38

109.73

109.73

109.74

109.74

1.44 3.13 16.05 0.09 1.41

2.01 3.13 16.06 0.13 1.41

2.15 3.13 16.07 0.13 1.41

2.87 3.14 16.10 0.18 1.41

0.00 109.73

0.00 109.73

0.00 109.74

0.00 109.74

4. Point 2 to Point 3 Average Flow 5Q/16 (m3/s) Flow depth WSEL @ 2 invert (106.60 m) (m) Flow area 5.13 m width depth (m3) Velocity flow/area (m/s) r A/P / (P w 2d) (m) Conduit loss [(V n)/(rr2/3)]2 L (m) where Manning’s n 0.014 and L 14.63 m WSEL at Point 3 (m)

0.68 3.13 16.05 0.04 1.41

0.96 3.13 16.06 0.06 1.41

1.03 3.13 16.07 0.06 1.41

1.37 3.14 16.10 0.08 1.41

0.00 109.73

0.00 109.73

0.00 109.74

0.00 109.74

5. Point 3 to Point 4 Average flow Q/8 (m3/s) Flow depth WSEL @ 3—invert (106.60 m) (m) Flow area 5.13 m width depth (m3)

0.27 3.13 16.05

0.38 3.13 16.06

0.41 3.13 16.07

0.55 3.14 16.10



22.22

Chapter Twenty-Two

TABLE 22.5

(Continued) Initial Operation Parameter

Velocity flow/area (m/s) r = A/P / (P w 2d) (m) Conduit loss [(V n)/(rr2/3)] 2 L (m) where n 0.014 and L 21.95 m WSEL at Point 4 (m) 6. Point 4 to Point 5 Flow Q/32 (m3/s) Port area 0.30 m deep 0.76 m wide (m2) Velocity flow/area (m/s) Submerged entrance loss 0.8 V2/2g (m) WSEL at Point 5 (in Sedimentation Tank) (m)

Design Operation

Min. Day. Avg. Day Avg. Day Max. Hour 0.02 1.41

0.02 1.41

0.03 1.41

0.03 1.41

0.00 109.73

0.00 109.73

0.00 109.74

0.00 109.74

0.07 0.23 0.29 0.00 109.73

0.10 0.23 0.41 0.01 109.74

0.10 0.23 0.44 0.01 109.74

0.14 0.23 0.59 0.01 109.76

23.16 0.77 105.97 3.77 87.36

23.16 0.82 105.97 3.77 87.41

23.16 1.09 105.97 3.79 87.68

9.71 0.01

9.71 0.01

9.74 0.01

7. Point 5 to Point 6 Width of sedimentation basin (W) W (m) 23.16 Flow (Q/4) (m3/s) 0.55 Invert elevation of sedimentation baffles (m) 105.97 Flow depth (H) H (WSEL at Point 5—baffle invert) (m) 3.76 Area downstreams of baffle (W H H) (m2) 87.21 Horizontal openings in baffle, 2.54 cm wide spaced every 22.86 cm. Area of openings A W .0254 H/.2286 (m2) 9.69 Velocity of downstream baffle (V downstream) 0.01 (Q/A) (m/s) Velocity of 2.54 cm opening section (V1) (Q/A / ) (m/s) 0.06 Local losses sudden expansion (1.0 (V downstream)2/2g) 2 sudden contraction (0.36 (VI) V / 2g) (m) 0.00 WSEL at Point 6 (Upstream of sedimentation baffles) (m) 109.73

0.08

0.08

0.11

0.00 109.74

0.00 109.74

0.00 109.76

8. Point 6 to Point 7 Loss per stage (provided by flocculator manufacturer) (m) 0.01 Total loss (three stages) (m) 0.04 WSEL at Point 7 (m) 109.77

0.01 0.04 109.78

0.03 0.09 109.83

0.05 0.15 109.91

9. Point 7 to Point 8 Flow Q/24 (m3/s) Port area 0.30 m deep 0.46 m wide (m2) Velocity flow / area (m/s) Entrance loss 1.25 V2/2g (m) WSEL at Point 8 (inlet port) (m)

0.09 0.14 0.65 0.03 109.80

0.13 0.14 0.92 0.05 109.83

0.14 0.14 0.98 0.06 109.89

0.18 0.14 1.31 0.11 110.02

0.09 0.68 0.62 0.15 0.27

0.13 0.72 0.65 0.19 0.28

0.14 0.77 0.71 0.19 0.29

0.18 0.90 0.82 0.22 0.30

0.00 109.80

0.00 109.83

0.00 109.89

0.00 110.02

0.18

0.26

0.27

0.36

Note: For Points 8 through 14, see Fig. 22.13 10. Point 8 to Point 9 Average flow Q/24 (m3/s) Flow depth WSEL @ 8 – invert (109.12 m) (m) Flow area 0.91 m width depth (m2) Velocity flow/area (m/s) r = A/P (P w 2d) (m) Conduit loss [(V n)/(rr2/3)] 2 L (m) where n 0.014 and L 3.86 m WSEL at Point 9 (m) 11. Point 9 to Point 10 Average flow Q/12 (m3/s)



Water and Wastewater Treatment Plant Hydraulics 22.23 TABLE 22.5


Design Operation

Min. Day. Avg. Day Avg. Day Max. Hour

Flow depth WSEL @ 9 – invert (109.12 m) (m) Flow area 0.91 m width depth (m2) Velocity flow/area (m/s) r = A/P / (P w 2d) (m) Conduit loss [(V n)/(rr2/3)]2 L (m) where n 0.014 and L 3.86 m WSEL at Point 10 (m)

0.68 0.62 0.29 0.27

0.72 0.65 0.39 0.28

0.77 0.71 0.39 0.29

0.90 0.82 0.44 0.30

0.00 109.80

0.00 109.84

0.00 109.89

0.00 110.02

12. Point 10 to Point 11 Flow Q/8, m3/s Flow depth WSEL @ 10 invert (109.12 m) (m) Flow area 0.91 width depth (m2) Velocity flow/area (m/s) 2 Loss at two 45° bends 2 0.2 V /2g (m) WSEL at Point 11 (m)

0.27 97.34 89.01 0.00 0.00 109.80

0.38 97.38 89.04 0.00 0.00 109.84

0.41 97.44 89.09 0.00 0.00 109.89

0.55 97.56 89.21 0.01 0.00 110.02

13. Point 11 to Point 12 Flow Q/4 (m3/s) Flow depth WSEL @ 11 invert (109.12 m) (m) Flow area 1.52 m width depth (m2) Velocity flow/area (m/s) Loss at two 45° bends 2 0.2 V 2/2g (m) r = A/P / (P w 2d) (m) Conduit loss [(V n)/(rr2/3)]2 L (m) where n 0.014 and L 9.75 m WSEL at Point 12 (m)

0.55 0.68 1.04 0.52 0.00 0.36

0.77 0.72 1.09 0.70 0.00 0.37

0.82 0.78 1.18 0.69 0.00 0.38

1.09 0.90 1.37 0.80 0.00 0.41

0.00 109.81

0.00 109.84

0.00 109.90

0.00 110.03

14. Point 12 to Point 13 Flow Q/4, (m3/s) Flow depth WSEL @ 12 invert (109.12 m) (m) Inlet area 1.52 m width depth (m2) Velocity flow/area (m/s) Inlet loss 1 V 2/2g (m) WSEL at Point 13 (Mixing Chamber No. 2 outlet) (m)

0.55 0.69 1.05 0.52 0.01 109.82

0.77 0.72 1.10 0.69 0.02 109.87

0.82 0.78 1.19 0.69 0.02 109.92

1.09 0.91 1.38 0.79 0.03 110.06

15. Point 13 to Point 14 Note: Mixers provide negligible head loss 0.55 Flow Q/4 (m3/s) Chamber area 1.83 m 1.83 m (m2) 3.34 Velocity flow/area (m/s) 0.16 Losses Mixer (1 V 2/2g) Sharp bend (1.8 V 2/2g) (m) 0.00 WSEL at Point 14 (Mixing Chamber No. 2 inlet) (m) 109.82

0.77 3.34 0.23 0.01 109.87

0.82 3.34 0.25 0.01 109.93

1.09 3.34 0.33 0.02 110.07

1.53 2.79 0.55 0.40

1.64 2.79 0.59 0.40

2.19 2.79 0.78 0.40

0.01

0.01

0.02

0.01 109.89

0.01 109.95

0.02 110.11

Note: For Points 14 through 21, see Fig. 22.14 16. Point 14 to Point 15 1.09 Flow Q/2 (m3/s) Conduit area 2.29 m wide 1.22 m deep (m2) 2.79 Velocity flow/area ( m/s) 0.39 R = A/P / (P 2w 2d) (m) 0.40 Conduit losses L [V/(0.849 V C R0.63)] 1/0.54 (m) where L 47.24 m and Hazen-Williams C 120 0.00 Local losses flow split (0.6 V 2/2g) contraction 0.01 (0.07 V 2/2g) 0.67 V 2/2g (m) WSEL at Point 15 (at Mixing Chamber No. 1) (m) 109.83



22.24

Chapter Twenty-Two

TABLE 22.5


17. The above calculations (for Points 1 through 15) have been for flow routed through Tank No. 4. When the flow is routed through Tank No. 1. the WSEL (m) is: In reality, the headloss through each basin is equal. The flow through the basin naturally adusts to equalize headlosses, i. e. flow through Tank No. 1 is greater than Q/4 and flow through Tank No. 4 is less than Q/4. The actual headloss through each basin can be estimated as the average of: Losses through Tank No’s. 1 and 4 and the WSEL (m) at Point 15 is: 18. Point 15 to Point 16 Flow Q (m3/s) Conduit area 2.29 m wide 1.22 m deep (m2) Velocity flow/area (m/s) R A/P / (P 2w 2d) (m) Conduit losses L [V/(0.849 V C R0.63)] 1/0.54 (m) where L 125.58 m and Hazen-Williams C 120 WSEL at Point 16 (m) 19. Point 16 to Point 17 Flow Q (m3/s) Conduit area @ 16 2.29 m wide 1.22 m deep (m2) Conduit area @ 17 1.68 m wide 1.68 m deep (m2) Average area (m2) Velocity flow / Area (m/s) R @ 16 A16/ (2 (2.29 m 1.22 m)) (m) R @ 17 A17/ (2 (1.68 m 1.68 m)) (m) Average R, (m) Conduit losses L [V/(0.849 V C R0.63)]1/0.54 (m) where L 9.14 m and Hazen-Williams C 120 WSEL at Point 17 (m) 20. Point 17 to Point 18 Flow Q (m3/s) Conduit area @ 17 1.68 m wide 1.68 m deep (m2) Velocity 17 flow/area 17 (m/s) 2 Pipe area @ 18 (D) (m) where D 1.68 m 4 Velocity 18 flow/area 18 (m) Exit losses V182/2g – V172/2g (m/s) WSEL at Point 18 (m) 21. Point 18 to Point 19 R = A/P / (P d ) (m) Local losses 3 elbows (3 0.25V 2/2g) entrance (0.5 V 2/2g) 1.25 V 2/2g (m) Conduit losses L [V/(0.849 V C R0.63)]1/0.54 (m) where L 138.68 m and Hazen-Williams C 120 WSEL at Point 19 (exit of Control Chamber) (m)

Design Operation


109.82

109.88

109.94

110.08

109.83

109.89

109.95

110.10

2.19 2.79 0.78 0.40

3.06 2.79 1.10 0.40

3.28 2.79 1.18 0.40

4.38 2.79 1.57 0.40

0.04 109.87

0.08 109.97

0.10 110.04

0.16 110.26

2.19 2.79 2.81 2.80 0.78 0.40 0.42 0.41

3.06 2.79 2.81 2.80 1.09 0.40 0.42 0.41

3.28 2.79 2.81 2.80 1.17 0.40 0.42 0.41

4.38 2.79 2.81 2.80 1.56 0.40 0.42 0.41

0.00 109.88

0.01 109.98

0.01 110.05

0.01 110.27

2.19

3.06

3.28

4.38

2.81 0.78 2.21 0.99 0.02 109.90

2.81 1.09 2.21 1.39 0.04 110.01

2.81 1.17 2.21 1.49 0.04 110.09

2.81 1.56 2.21 1.98 0.8 110.35

0.42

0.42

0.42

0.42

0.06

0.12

0.14

0.25

0.07 110.03

0.13 110.27

0.15 110.39

0.26 110.86





22. Point 19 to Point 20 Weir elevation (m) Depth of flow over weir (WSEL @ 19 – weir elevation), (m) Length of weir, L, (m) 3/2 0.385 Flow over weir q 1.71 h3/2 [ 1 (d / n) ] L Note: Rather than solve for h, find an h by trial and error that gives a q equal to the flow for the given flow scenarios (given in Item 1) assume h (m) then q (m3/s) assume h (m) then q (m3/s) Note: These q’s equal the flows for the given scerios (Item 1) h (m) WSEL at Point 20 (h WSEL @ Point 19) (m) 23. Point 20 to Point 21 Flow Q (m3/s) Sluice gate area 1.37 m 1.37 m (m2) Velocity Flow/Area (m/s) Gate Losses 1.5 V 2/2g (m) WSEL at Point 21 (Raw Water Control Chamber) (m) The overflow weir in the Raw Water Control Chamber is 3.05 m long and is sharp crested Q = 1.82 L h3/2 so h (Q/1.82L)2/3 (m) The water surface must not rise above elevation 112.78 m The overflow weir elevation may be safely set at 111.86 m

Design Operation


109.73

109.73

109.73

109.73

0.30 2.74

0.54 2.74

0.66 2.74

1.13 2.74

0.60 1.84 0.66 2.18

0.90 3.14 0.89 3.07

0.95 3.12 0.97 3.27

1.35 4.21 1.37 4.42

0.66 110.39

0.89 110.62

0.97 110.70

1.37 111.10

2.19 1.88 1.16 0.10

3.06 1.88 1.63 0.20

3.28 1.88 1.74 0.23

4.38 1.88 2.33 0.41

110.49

110.82

110.93

111.51

0.54

0.67

0.70

0.85

hydraulic reaches analyzed in the example. The circled numbers indicate points at which the WSEL is calculated. Hydraulic calculations start downstream of the sedimentation basins (Fig. 22.12) and proceed upstream through the mixing chamber (Fig. 22.13) and the Raw Water Control Chamber (Fig. 22.14). Mechanical mixers and mechanical flocculators are used. Conduit losses between the rapid mix chambers and the Raw Water Control Chamber are also calculated in the example. Three different flow rates (i.e., minimum day, average day, and, maximum hour) are used in the calculations. This is a range of design flow conditions that a design engineer would typically take into consideration. The longest path through the flocculation and sedimentation processes, through Basin No. 4, is followed (Points 1 through 15). Although not shown, losses along the shortest path have also been calculated. As would be expected, the calculated head loss is smaller for the shorter path. The actual losses are equal for each path. The flows through each path naturally adjust to equalize losses. The flow through the longest path is slightly smaller than the flow through the shortest path. In the example, the WSEL at Point 15 is adjusted to reflect the average losses through the basins. The WSEL calculations upstream of Point



22.26

Chapter Twenty-Two

FIGURE 22.12 Flocculation/ sedimentation basin

15 are based on the adjusted WSEL. Alternatively the weirs or ports feeding flow into each basin may be adjusted to create an equal distribution of flows in all basins as discussed in Sec. 22.2.1. 22.3.7.2.2 Filtration. Process criteria. Suspended solids are removed from the water as it passes through a porous medium during filtration. Filters operate under either gravity or pressure. Filters also differ in the type and distribution of the media used (fine, course, uniformly graded, graded coarse to fine, etc.) and the direction of flow through the media (upflow, downflow, and biflow). Pressure filter hydraulics information is very product specific and should be obtained from the manufacturer. The design engineer using pressure filters should then apply this information to the project using project–specific hydraulic considerations. This section presents information on gravity filters. Key hydraulic design parameters. The headloss through a filter increases with use as the voids become filled with solid particles. When the headloss reaches a certain point (terminal headloss), the filter is backwashed to remove the solids. The rate of headloss buildup is dependent on several factors, including how the filter is graded (the arrangement of media particle sizes). The rate of headloss buildup is reduced (and filtration is more effective) when the flow first goes through the coarse media and then the fine media.




FIGURE 22.13 Mixing chamber

FIGURE 22.14 Raw water control chamber



22.28

Chapter Twenty-Two

However, during backwash, the high rate of flow expands the filter bed and, over time, the media are regraded so that the more coarsely graded grains are located at the bottom and the fines are located at the top. To benefit from the coarse-to-fine grading, an upward flow pattern can be used, but is very uncommon. More often the filter media are selected such that the fine media have a higher specific gravity than the coarse media to maintain the course-to-fine gradation during backwash. The most commonly used filter media are natural silica sand and crushed anthracite coal; however garnet and ilmenite are used in mixed media beds. Granular carbon is often used if taste and odor control is desired. The terminal headloss is determined by a combination of factors including filter breakthrough (when the filter bed loses its adsorptive capacity), available static head, and outlet pressure required. The filter should be designed so that the headloss in any level of the filter bed does not exceed the static pressure. A negative head can result in air binding in the filter which will, in turn, further increase headloss. Filter influent piping is sized to limit velocities to about (0.6 m/s). Wash-water and effluent piping flow velocities are kept below (1.8 m/s) so that hydraulic transients(waterhammer) and excessive headlosses are minimized and controlled to within tolerable limits. Hydraulic design example. Table 22.6 illustrates the calculation of the WSEL from the clear well back upstream to the Sedimentation Basin effluent at the medium-sized water treatment plant shown in Fig. 22.10. Figures 22.15 and 22.16 show details of the hydraulic reaches analyzed in the example. Table 22.7 illustrates the filter hydraulic calculation, the details of which are shown in Figs. 22.17 and 22.18. The hydraulic profile of the plant (based on hydraulic calculations done in Tables 22.5, 22.6 and 22.7) is shown in Figure 22.3.

TABLE 22.6 Hydraulic Calculations in a Medium–Sized Water Treatment Plant from the Filter Effluent to the Effluent Clearwell Initial Operation Parameter 1. Flow (m3s) Note: for Points 22 through 28, see Figure 22.15 2. Point 22 to Point 23 Maximum water level in Clearwell (Point 22) (m) Invert in Clearwell (m) Flow Q/2 (m3/s) Stop logs @ A Flow area (2 openings, 1.52 m wide, 3.66 m deep) (m2) Velocity flow/area (m/s) Loss 0.5 V 2/2g (m) Baffles Flow area (3.05 m wide, 3.66 m deep) (m2) Velocity flow/area (m/s) Loss 1.0 V 2/2g (m) Stop logs @ B and C Same as the losses @ A, times 2 (m)

Min Day

Avg Day

Design Operation Avg Day Max Hour

2.19

3.06

3.28

4.38

105.16 101.50 1.09

105.16 101.50 1.53

105.16 101.50 1.64

105.16 101.50 2.19

11.15 0.20 0.00

11.15 0.27 0.00

11.15 0.29 0.00

11.15 0.39 0.00

11.15 0.20 0.00

11.15 0.27 0.00

11.15 0.29 0.00

11.15 0.39 0.01

0.00

0.00

0.00

0.01





WSEL at Point 23 (m)

Min Day

Avg Day


105.16

105.17

105.17

105.18

1.09

1.53

1.64

2.19

2.21 0.50 0.01 0.01 0.01

2.21 0.69 0.02 0.01 0.01

2.21 0.74 0.03 0.01 0.01

2.21 0.99 0.05 0.03 0.03

0.00 105.19

0.00 105.22

0.00 105.23

0.00 105.28

0.55 2.32 0.24 0.00

0.77 2.32 0.33 0.01

0.82 2.32 0.35 0.01

1.09 2.32 0.47 0.01

0.00

0.00

0.00

0.00

5. Point 25 to Point 26 Sluice Gate No. 1 flow area 1.22 m 0.91 m (m2) Velocity Q/A / (m/s) Loss 0.5 V 2/2g (m) WSEL at Point 26 (m)

1.11 0.49 0.01 105.20

1.11 0.69 0.01 105.24

1.11 0.74 0.01 105.24

1.11 0.98 0.02 105.32

6. Point 26 to Point 27 Sluice Gate No. 2 Loss 0.8 V 2/2g (m) WSEL at Point 27 (m)

0.01 105.21

0.02 105.25

0.02 105.27

0.04 105.36

3. Point 23 to Point 24 Flow Q/2 (m3/s) 1.68 (m) diameter pipe Flow area d 2/4 (m2) Velocity flow/area (m/s) Exit loss @ clearwell V 2/2g (m) o Loss @ 2 - 90 bends (0.25 V 2/2g) 2 (m) Entrance loss @ Filter Building 0.5 V 2/2g (m) Pipe loss (3.022 V 1.85 L)/ (C 1.85 D 1.165 ) where C 120 and L 57.91 m (m) WSEL at Point 24 (m) 4. Point 24 to Point 25 Flow Q/4 (m3/s) Flow area 1.52 m 1.52 m2 Velocity Q/A (m/s) Loss as flows merge 1.0 V 2/2g (m) Conduit loss [(V n)/(R 2/3 )]2 L (m) where n 0.013, L 16.76 m and R A/P / (P 6.10 m) WSEL at Point 25 (m)

7. Point 27 to Point 28 Port to Filter Clearwell: Calculate losses through port as if were a weir when depth of flow is below top of port. Port dimmensions 2.74 m wide by 0.813 m deep. Flow Q/4 (m3s) Weir (bottom of port) elevation (m) Depth of flow over weir (WSEL @ 27 – weir elevation) (m) Flow over submergedweir q 1.71 h3/2 [1 - (d/ d h)3/2]0.385 L Note: Rather than solve for h, find an h, by trial and error, that gives a q equal to the flow for the given flow scenario assume h (m) then q (m3/s) assume h (m) then q (m3/s) Note: These q’s equal the flows for the given scenarios h (m)

0.55 104.85

0.77 104.85

0.82 104.85

1.09 104.85

0.36

0.40

0.42

0.51

0.40 0.59 0.39 0.52

0.45 0.69 0.46 0.76

0.50 0.95 0.48 0.82

0.60 1.23 0.58 1.09

0.39

0.46

0.48

0.58



22.30

Chapter Twenty-Two

TABLE 22.6


WSEL at Point 28 (m) Filters—See Filter Hydraulics in Table 22.7 Note: for Points 29 through 33, see Fig. 22.16 8. Point 29 WSEL above filters (m) 9. Point 29 to Point 30 Entrance to Filter #4 Flow, Q/8 (m3/s) Channel velocity = flow/area (area 1.22 m 1.22 m) (m/s) Submerged entrance loss 0.8 V 2/2g (m) 1.22 m pipe velocity flow/area (area d 2/4 ) (m/s) Butterfly valve loss 0.25 V 2/2g (m) Sudden enlargement loss 0.25 V 2/2g (m) WSEL in influent channel (Point 30) (m) 10. Point 30 to Point 31 Flow depth WSEL @ 30 invert (107.29 m) (m) Flow area 1.83 m width depth (m2) Velocity flow/area (m/s) R = A/P / (P w 2d) (m) Conduit Loss [(V n)/(rr2/3)]2 L where n 0.014 and L 10.77 m (m) WSEL at Point 31 (m) 11. Point 31 to Point 32 Flow Q/4 (m3/s) Flow depth WSEL @ 31 - invert (107.29 m) (m) Flow area 1.83 m width depth (m2) Velocity flow/area (m/s) R = A/P / (P w 2d) (m) Conduit loss [(V n)/(r 2/3 )]2 L (m) where n 0.014 and L 10.77 m WSEL at Point 32 (m) 12. Point 32 to Point 33 Flow 3Q/8 (m3/s) Flow depth WSEL @ 32 – invert (107.29 m) (m) Flow area 1.83 m width depth (m2) Velocity flow/area (m/s) R A/P / (P w 2d) (m) Conduit loss [(V n)/(rr2/3)]2 L (m) where n 0.014 and L 10.77 m WSEL at Point 33 (m) 13. Point 33 to Point 1 Flow Q/2 (m3/s) Flow depth WSEL @ 33 – invert (107.29 m) (m) Flow area 1.83 m width depth (m2) Velocity flow/area (m/s) R = A/P / (P w 2d) (m) Conduit loss [(V n)/(r 2/3 )]2 L (m) where n 0.014 and L 11.07 m WSEL at Point 1 (m)

Design Operation

Min Day

Avg Day

Avg Day Max Hour

105.24

105.31

105.33

105.43

109.73

109.73

109.73

109.73

0.27

0.38

0.41

0.55

0.18 0.00

0.26 0.00

0.28 0.00

0.37 0.01

0.23 0.00 0.00 109.73

0.33 0.00 0.00 109.73

0.35 0.00 0.00 109.73

0.47 0.00 0.00 109.74

2.44 4.46 0.06 0.67

2.44 4.47 0.09 0.67

2.44 4.47 0.09 0.67

2.45 4.48 0.12 0.67

0.00 109.73

0.00 109.73

0.00 109.73

0.00 109.74

0.55 2.44 4.46 0.12 0.67

0.77 2.44 4.47 0.17 0.67

0.82 2.44 4.47 0.18 0.67

1.09 2.45 4.48 0.24 0.67

0.00 109.73

0.00 109.73

0.00 109.73

0.00 109.74

0.82 2.44 4.46 0.18 0.67

1.15 2.44 4.47 0.26 0.67

1.23 2.44 4.47 0.28 0.67

1.64 2.45 4.48 0.37 0.67

0.00 109.73

0.00 109.73

0.00 109.73

0.00 109.74

1.09 2.44 4.46 0.24 0.67

1.53 2.44 4.47 0.34 0.67

1.64 2.45 4.47 0.37 0.67

2.19 2.45 4.48 0.49 0.67

0.00 109.73

0.00 109.73

0.00 109.74

0.00 109.74




FIGURE 22.15 Clearwell to filter effluent

FIGURE 22.16 Filter effluent



22.32

Chapter Twenty-Two

TABLE 22.7

Example Hydraulic Calculation of a Typical Filter

Parameter Plant flow (m3/s) Filter loading, [(m ⴢ m)/m ] Filter area per filter—seven (7) out of eight (8) filters in operation (m2) Flow loading area (m3/s) Losses through filter effluent piping (Fig. 22.17) 0.51 m piping (Q): Pipe velocity Q/A / (m/s) Local losses Exit (0.5) butterfly o valves (2 0.25) 90 elbows (2 0.4) 2 tee (1.8) 3.6 V /2g (m) R A/P / (d 2/4 p)/(d p) dd/4 (m) Conduit losses L [V/(0.849 V C R0.63)] 1/0.54 where L 6.10 m and HazenWilliams C 120 (m) 0.51 m piping (Q/2): Pipe velocity Q/A (m/s) Local Losses Butterfly Valve (0.25) (m) R A/P / (d 2/4 p)/(d p) dd/4 (m) Conduit losses L [V/(0.849 V C R0.63)] 1/0.54 where L 3.05 m and HazenWilliams C 120 (m) 0.61 m piping (Q/2): Pipe velocity Q/A / (m/s) Local losses entrance (1.0) tee (1.8) 2.8 V 2/2g (m) Filter (clean) and underdrain losses (obtain from manufacturer) (m) Total losses (effluent pipe and clean filters) (m) 3

2

Initial Operation Min. Day Avg. Day.

Design Operation Avg. Day. Max. Hour.

2.19

3.06

3.28

4.38

0.083 115

0.167 115

0.250 115

0.334 115

0.16

0.32

0.48

0.64

0.79

1.58

2.37

3.16

0.11 0.13

0.46 0.13

1.03 0.13

1.83 0.13

0.01

0.03

0.06

0.11

0.40 0.00 0.13

0.79 0.01 0.13

1.19 0.02 0.13

1.58 0.03 0.13

0.00

0.00

0.01

0.02

0.27

0.55

0.82

1.10

0.01

0.04

0.10

0.17

0.09 0.23

0.15 0.70

0.23 1.45

0.34 2.50

Assume that headloss will be allowed to increase 2.44 m before the filters are backwashed. A rate controller will be used to maintain a constant flow through the filters. Determine the ranges of available head over which the rate controller will operate. Static Head (Fig. 22.18) WSEL above filters (m) WSEL in filter effluent conduit, Point 29 (see Example 22.2) break Maximum (m) Minimum (m) Static head WSEL above filters—WSEL at Point 29 (Filter effluent conduit-2) Maximum (m) Minimum (m) Available head static head 2.44 m Maximum (m) Minimum (m)

109.73

109.73

109.73

109.73

105.61 105.16

105.61 105.16

105.61 105.16

105.61 105.16

4.57 4.11

4.57 4.11

4.57 4.11

4.57 4.11

2.13 1.68

2.13 1.68

2.13 1.68

2.13 1.68




Head, meters

FIGURE 22.17 Filter effluent piping

FIGURE 22.18 Available head over which filter effluent rate controller operates—metric units.



22.34

22.3.8

Chapter Twenty-Two

Membrane Technology

Membranes are synthetic filtering media manufactured from a variety of materials including polypropylene, polyamide, polysulfone, and cellulose acetate. The membrane material can be arranged in various configurations, including the following: • Spiral wound • Hollow fiber • Tubular • Plate frame Examples of these configurations are presented in Fig. 22.19. In water and wastewater treatment applications, the most common configurations are spiral wound and hollow fiber. In general, there are four classes of membranes: microfilters (MF), ultrafilters (UF), nanofilters (NF), and hyperfilters. Treatment through hyperfilters is referred to as hyperfiltration, or reverse osmosis (RO). The hydraulics associated with membranes are membrane-specific and can be obtained from the manufacturer. This section presents general considerations pertinent to flow through membranes. As with natural particle media filters, clean membranes have a specific headloss and, over time, as the membranes become covered with a cake buildup, the effectiveness of the membrane decreases and headloss increases. Fouling (excessive buildup) may damage the membrane. The need for pretreatment ahead of membranes is determined by the raw water quality and the membrane type. In general, microfilters and ultrafilters do not require pretreatment for treating surface or groundwater. Nanofilters and reverse osmosis membranes may require pretreatment depending on the type of fouling. Membrane fouling can result from particulate blocking, chemical scaling, and biological growth within the membranes. An estimate of particulate blocking can be made using indices such as the Silt Density Index (SDI) and the Modified Fouling Index (MFI). These fouling indices are determined from simple bench membrane tests using 0.45 micron Millipore filters and monitoring flow through the filter at a given pressure, usually 30 psig. Approximate values of suitable SDIs for nanofiltration are 0–3 units, and for reverse osmosis, 0–2 units. Corresponding values of MFI are, for nanofiltration 0 to 10 s/L2, and for RO, 0–2 s/L2. Scaling control is essential in RO and nanofilter membrane filtration, especially when the filtration provides water softening. Controlling precipitation or scaling within the membrane element requires identification of limiting salt, acid addition for prevention of calcium carbonate precipitation within the membrane, and/or the addition of an antiscalant. The amount of antiscalant or acid addition is determined by the limiting salt. A diffusion controlled membrane process will naturally concentrate salts on the feed side of the membrane. As water is passed through the membrane, this concentration process will continue until a salt precipitates and scaling occurs. Scaling will reduce membrane productivity and, consequently, recovery is limited by the allowable recovery just before the limiting salt precipitates. The limiting salt can be determined from the solubility products of potential limiting salts and the actual feed stream water quality. Ionic strength must also be considered in these calculations as the natural concentration of the feed stream during the membrane process increases the ionic strength, allowable solubility and recovery. Calcium carbonate scaling is commonly controlled by sulfuric acid addition, although sulfate salts, such as barium sulfate and strontium sulfate, are often the limiting salt. Commercially available antiscalants can be used to control scaling by complexing the




FIGURE 22.19 Membrane configurations. (a) Spiral wound, (b) hollow fiber, (c) tubular, (d) plate and frame.

metal ions in the feed stream and preventing precipitation. Equilibrium constants for these antiscalants are not available which prohibits direct calculation. However, some manufacturers provide computer programs for estimating the required antiscalant dose for a given recovery, water quality, and membrane. Biological fouling is controlled with some membranes such, as cellulose acetate, by maintaining a free chlorine residual of not more than 1 mg/L. Other membranes, such as the thin-film composites, are not chlorine tolerant and must rely on upstream disinfection by, for example, ultraviolet disinfection or chlorination-dechlorination. The extent of fouling for a specific application and its influence in the design of nanofiltration and RO membrane systems is best determined by pilot studies. It has been suggested that some buildup on the membrane may be beneficial to treatment by providing an additional filtering layer. At facilities operated by the Metropolitan Water District of Southern California (MWD), removal rates of 1.7–2.9 logs were



22.36

Chapter Twenty-Two

observed for seeded virus MS2 bacteriophage through microfilters that had a pore size an order of magnitude larger than the nominal size of MS2 (1). The microfiltration system used by MWD utilizes an air backwash procedure whereby compressed air at 90–100 psig is introduced into the filtrate side of the hollow fiber membranes. Accumulated particulates dislodged by the compressed air are swept away by raw water introduced to the feed side of the membranes. The backwash sequence is carried out automatically at preset time intervals. MWD found the best interval to be every 18 minutes. The total volume of backwash represents approximately 5–7 percent of influent flow. The difference between influent and effluent pressure across the membrane is termed the transmembrane pressure (TMP). Despite the frequent air and water backwashes, the TMP gradually increases over time. Generally, when the TMP reaches approximately 15 psig, chemical cleaning of the membranes is carried out. If the TMP is allowed to increase beyond 15 psig, particulates can become deeply lodged within the lattice structure of the membranes and will not be removed, even by chemical cleaning. Chemical cleaning typically lasts 2–3 hours and involves circulating a solution of sodium hydroxide and a surfactant through the membranes, and soaking them in the solution. The membranes at the MWD microfilter plants have a surface loading rate of 40–67 ft2. The lower flux rate of 40 ft2 has the advantage that the rate of increase of TMP is reduced and the interval between chemical cleanings is increased. A possible explanation for this is that particulates are not forced as deeply into the lattice structure of the membranes, thereby allowing the air-water backwash to clean the membranes more effectively. By reducing the flux rate from 67–40 ft2, the interval between chemical cleanings was increased from 2 to 3 weeks to almost 20 weeks. However, MWD has instituted a maximum run time of 3 months between chemical cleanings to ensure the long-term integrity of the membranes. Nanofiltration is widely used for softening groundwaters in Florida. A typical nanofiltration plant would include antiscalant for scale control added to the raw water. Cartridge filters, usually rated at 5 microns, remove particles that may foul the membrane system. Feed water pumps boost the pretreated water pressure to about 90–130 pounds per square inch (psi) before entering the membrane system. The membranes typically are spiral wound nanofiltration membranes generally with molecular weight cutoff values in the 200–500 dalton range.

22.4 WASTEWATER TREATMENT Many factors and considerations influence the hydraulic design of a wastewater treatment plant. This section describes typical phases of wastewater treatment planning required for design of new plants or additions to existing plants, and then presents typical unit process hydraulic computations.

22.4.1 Wastewater Treatment Planning Hydraulic decision making for a new wastewater treatment plant or expansion of an existing plant involves several planning phases. Typical planning phases are presented below in their common order of consideration. 22.4.1.1 Service area and flows. More than 15,000 municipal wastewater treatment plants are in operation in the United States today. The plants are designed to treat a total of about




140 million m3 of flow each day. Flow quantities requiring treatment change over time based on a number of factors related to service area. These factors include the following: Changes in service area size. Most often the service area will increase in size during the wastewater treatment plant service life. However, service area size may decrease, such as when wastewater in larger metropolitan areas is diverted to an alternate wastewater treatment plant. Information about anticipated changes in the size of a wastewater treatment plant service area can sometimes be found in “regional planning” documents. Changes in service area land use. Changes in the type of land use in the service area, such as from residential to industrial, will impact the flow rates to be served by the treatment plant. Also, the development of impervious areas within the wastewater treatment plant service area will reduce infiltration and increase runoff volume and rate. If this runoff then enters the sewer system it will impact the flow rate to the plant. A combined sewer system will be more susceptible to this type of change than a separate sewer system. Changes in service area density. Wastewater treatment plant flows are a function of the number of inhabitants and industries which generate the wastewater. An understanding of the regional planning issues which may affect the wastewater treatment plant service area assists in estimating future increases in flow and making appropriate provisions for future plant expansions. Such flow increases will likely be partially offset by increased water conservation in water-limited areas. Changes in service area infiltration/inflow. Most often the rates of infiltration/inflow will increase as the collection system becomes older. Such flow increases can generally be offset by periodic sewer rehabilitation, manhole rehabilitation, and enforcement of inflow control ordinances. The quantity of wastewater to be handled by a wastewater treatment plant is affected primarily by the type of wastewater produced in the service area and type of wastewater collection system used. The four types of wastewater which may be produced in a given sewer system service area include sanitary wastewater, industrial wastewater, stormwater, and infiltration/inflow. The three types of sewer systems used to collect some or all of these flow types include sanitary, storm, and combined-sewer systems. The types of wastewater are defined as follows: Sanitary flow. Wastewater discharged from residences and from institutional, commercial and similar facilities. Quantities of sanitary flow can be estimated on a per capita basis for each type and size of residence or facility producing the flow. Industrial flow. Wastewater discharged from industrial facilities. In a heavily industrialized area, industrial flow can make up a majority of a wastewater plant’s influent flow. Industrial wastewater quantities produced by a given facility can be estimated based on facility type, size, and rate of production. Stormwater. Stormwater is precipitation runoff. Stormwater enters storm or combined collection systems as surface or subsurface inflow. The rate of stormwater entering a storm or combined sewer system as inflow mirrors the intensity and quantity of the precipitation event, although if the precipitation is frozen the runoff will be delayed until melting occurs. Infiltration Water (including stormwater) that seeps into a wastewater collection system through the ground, usually through cracks or leaks in the collection system. Accordingly, infiltration rates typically vary both annually and seasonally. The age of the



22.38

Chapter Twenty-Two

collection system should be considered when estimating infiltration rates because older collection systems are prone to higher infiltration rates. If the amount of infiltration is significant enough to affect plant influent water quality, the treatment processes must be selected accordingly. Inflow. Surface and subsurface stormwater discharging directly into a wastewater collection system. Precipitation events significantly impact inflow rates and can also impact infiltration rates by surcharging the groundwater table. The elevation of the groundwater table relative to the sewer elevation directly affects the infiltration flow rate. The types of sewer systems include sanitary-sewer systems which collect sanitary wastewater, industrial wastewater (if present in service area), and infiltration/inflow. Storm-sewer systems collect stormwater and infiltration/inflow, and combined-sewer systems collect sanitary wastewater, industrial wastewater (if present in service area), stormwater, and infiltration/inflow. Flows to wastewater treatment plants are conveyed by separate-sewer systems and, in some older systems, combined-sewer systems. Hydraulic design guidelines for sanitary-sewer systems have been compiled by the American Society of Civil Engineers and the Water Environment Federation (1982). 22.4.1.2 Effluent requirements. Treated wastewater can be discharged to rivers, lakes, oceans, and groundwater. There is also increasing re-use of wastewater for nonpotable applications, such as irrigation and industrial processing. Effluent quality requirements for wastewater treatment plants are generally established by regulatory agencies in the plant’s NPDES permit. Those minimum acceptable effluent characteristics and the anticipated influent characteristics determine what level of treatment is required and, thereby, determines to some degree what treatment processes are needed. Because each process type requires a different amount of head, the influent characteristics and effluent requirements also indirectly affect the plant head requirements. 22.4.1.3 Process selection. Each unit process in a wastewater treatment plant flow train treats the wastewater physically, chemically or biologically, or in some combination thereof. Because various combinations of unit processes are generally available to produce the desired effluent quality, the designer must choose among the options to select the optimum combination. In anticipation of future requirements, potential changes in effluent requirements and corresponding treatment train modifications should also be considered. Typical unit treatment processes for new wastewater treatment plants include screening, grit removal, primary sedimentation, aeration, secondary sedimentation, granular media filtration, disinfection, dechlorination, and postaeration. Figure 22.20 is a flow diagram showing how the typical processes are interconnected. Unless only one treatment process combination is capable of adequately treating the wastewater, pertinent factors must be used to select the process train. Typical factors include capital and operating costs, environmental impacts, aesthetics, and public acceptance. Process head requirements can directly affect capital costs, as those processes with higher head requirements are more likely to necessitate costly pumping facilities and deep structure excavation. 22.4.1.4 Hydraulic bases for design. Flow rates for the wastewater treatment plant must be established for the hydraulic design. Design year flow projections are often based on estimated conditions 15–20 years in the future. Providing sufficient treatment capacity to accommodate new development can be an important municipal commodity for expanding the municipal tax base. The design should also provide allowances for the initial plant operation when flow may be significantly less than the design flow, as well as expansion or rehabilitation to handle flows reasonably anticipated beyond the design year.




Peak flow is used for hydraulic design, whereas average flow is used for treatment process design. Peak flow is defined as the maximum hour flow experienced by the wastewater treatment plant throughout its service life. The maximum hour flow is generally two to five times the average daily flow. Plants serving combined collection systems can experience even greater flow variations. Treatment plant unit processes must convey the maximum flow unless this flow would cause a hydraulic washout of the treatment plant. In this situation, the designer should consider the use of equalization basins to minimize negative impact on the treatment process. In addition, the plant must also be able to fully process minimum flow without undesirable settling of solids throughout the treatment train. Plants normally encounter diurnal fluctuation of pollutant loadings, as well as flow loadings. Fluctuation in pollutant loadings may impact treatment process selection and consequently impact process hydraulics. 22.4.1.5 Flow diagram. A flow diagram should be prepared to depict the results of process selection and hydraulic bases of design. Details in a flow diagram should include the type of unit processes, number of basins for process redundancy, flow distribution and junction chambers, piping, and conduits for interconnecting the unit processes and major recycle streams such as return-activated sludge (RAS). Figure 22.20, which was mentioned above, shows a typical flow diagram. 22.4.1.6 Plant siting. Several factors affect the plant site selection process, including site elevation, topography, geology, and hydrology; site access; utility availability; seismic activity; surrounding land use and future availability; noise, odor and air quality requirements at and near the site; existing collection system and receiving water proximity; and other environmental considerations. A site’s hydraulic suitability for a wastewater treatment plant is determined primarily by site elevation and topography. The typical site elevation is low-lying, which facilitates the flow of wastewater from the service area by gravity and minimizes costly pumping in the collection system. Such a site, however, may require flood protection. The difference in head between the plant influent sewer and the receiving water body is the head available for the treatment plant. If available head does not exceed the plant’s head requirements, additional head can be provided by pumping the wastewater. Selecting processes with lower head requirements can also reduce the need for pumping. Pumping of wastewater, especially untreated wastewater, should be avoided when possible due to potential operational difficulties of handling the associated rags, grit, stringy material and other large solids. A mild, continuous slope usually provides optimal gravity flow conditions. Relatively flat sites often necessitate higher pumping heads. Sites on a severe, uneven slope or slopes can require costly hydraulic and structural features, and should be avoided when possible. 22.4.1.7 Plant layout. The selected treatment processes establish the major space and hydraulic requirements needed to develop initial plant layouts. Also, provisions for future unit process additions and plant capacity expansions should be included both spatially and hydraulically. Support facilities, such as maintenance, laboratory and administrative facilities, must also be considered. Arranging process elevations to generally follow plant site topography minimizes the amount of structural excavation. Site geology constraints may limit the practical depth and elevation of the processes. In such cases, additional pumping facilities may be necessary to provide sufficient head for the required water surface elevation. When arranging treatment processes, a preliminary hydraulic profile should be developed as discussed below. The plant hydraulic profile and site topography and geology information together determine the location having the optimal elevation for each process.



Chapter Twenty-Two

FIGURE 22.20 Schematic flow diagram of typical wastewater treatment plant.

22.40




Other objectives when developing a plant layout at a selected site include: close proximity of processes to associated facilities; structure grouping according to process; transportation equipment and staff traffic pattern efficiency; minimization of process piping; and safe, isolated hazardous chemical and material locations. When preparing layouts for addition of a new process to an existing plant, the existing plant hydraulic profile should be consulted to determine the amount of head available for the new process. If adequate hydraulic head is not available for the new process, new pumping facilities will be necessary. 22.4.1.8 Hydraulic profile and calculations. A hydraulic profile should be prepared for the flow train to graphically depict the results of hydraulic calculations and site layouts. Details in a profile should include free water surface elevations throughout the flow train, including unit treatment processes, interconnecting piping and channels, junction chambers, flowmeters and flow control devices, as well as structural profiles. Figure 22.21 shows a typical hydraulic profile. Both high and low water levels are shown to illustrate the range of liquid levels anticipated at each structure. Sufficient freeboard must be provided to prevent liquid or floating material from splashing over the sides under conditions of high water level. Low water levels are important when designing devices requiring a mimimum amount of submergence, such as surface skimmers or baffles. In addition to normal high and low water levels, hydraulic calculations should address other potential conditions. For example, for each process having redundant structures, the largest capacity unit should be assumed to be out of service during maximum flow for consideration of a “worst case”. The process structure should always be hydraulically capable of accommodating the change in elevation due to the “worst case.” head requirements without liquid overtopping the walls. The process head requirement is the amount of head lost by the wastewater as it passes through a process at maximum flow. The head requirement for a specific process can vary with flow rate, influent water quality, process equipment size, process equipment layout, process equipment components included, and process equipment manufacturer.

22.4.2 Typical Unit Process Hydraulics 22.4.2.1 Bar screens. Process criteria. The first unit operation typically encountered in a wastewater treatment plant is screening. A schematic diagram of a typical bar screen system is shown in Fig. 22.22. A screen is comprised of a screening element with circular or rectangular openings designed to retain coarse sewage solids. The screens are designated as hand cleaned or mechanically cleaned based on the method of cleaning. Based on the size of the openings, screens are designated as coarse or fine. The general dividing line between coarse and fine screens is an opening size of 6 mm (1/4 in). A bar screen is a coarse screen designed to remove large solids or trash that could otherwise damage or interfere with the downstream operations of treatment equipment, such as pumps, valves, mechanical aerators, and biological filters. The bar screens are oriented vertically or at a slope varying from 30°– 80° with the horizontal. Key hydraulic design parameters. The key hydraulic design parameters for bar screens include the approach channel, effective bar opening, and operating head loss. Approach channel. Velocity distribution in the approach channel is an important factor in successful bar screen operation. A straight channel ahead of the channel provides good velocity distribution across the screen and promotes effectiveness of the device. Use of a configuration other than a straight approach channel has often resulted in uneven flow



Chapter Twenty-Two

FIGURE 22.21 Typical hydraulic profile for wastewater treatment plant.

22.42




distribution within the channel and accumulation of debris on one side of the screen. The velocity in the approach channel should be maintained at a self-cleaning value to dislodge deposits of grit or screenings. Ideally, the velocity in the screen chamber should exceed 0.4 m/s (1.3 ft/s) at minimum flows to avoid grit deposition if grit chambers follow bar screens. However, this is not always practical with the typical diurnal and seasonal fluctuation in wastewater flows. In general, common design practice provides velocities of 0.6–1.2 m/s (2–4 ft/s) for mechanically cleaned bar screens and 0.3–0.6 m/s (1–2 ft/s) with a velocity of 0.9 m/s (3 ft/s) at peak instantaneous velocity for manually cleaned bar screens. Effective bar opening. Various types of bar screens, including trash racks, manual screens and mechanically cleaned bar screens, employ a wide range of openings from 6 to 150 mm (14–6 in). The smaller screen openings collect larger quantities of screenings and generally produce higher head losses. The effective area of the screen openings equals the sum of the vertical projections of the screen openings. Operating head loss. As the screenings are collected, the openings in the screen become partially clogged and head losses increase. The maximum design allowance for headloss through the clogged screens is generally limited to 0.8 m (2.5 ft). Curves and tables for head loss through the screening device are usually available from the equipment manufacturer. To prevent flooding of the screening area caused by severe blinding of the screen during a power failure or similar disruption to cleaning, the design should provide for an overflow weir or gate and a parallel channel allowing overflows to flow around the screen. Hydraulic design example. The wastewater influent transported through the inlet sewer passes the bar screens prior to discharge into the pump well. Three bar screens are provided to handle hydraulic loadings varying from 1.0 m3/s (23 mgd) for minimum day flow during initial operation to 3.2 m3/s (73 mgd) for maximum hour flow during design operation. Sluice gates and stop logs are provided as part of the bar screen design so that any bar screen can be isolated for maintenance as required. Design hydraulic calculations for the bar screens are shown in Table 22.8. The WSEL at the pump well provides a downstream control point for the bar screens and channels. The WSEL at the pump well normally fluctuates between the pump control high water level and low water level. A high water level (HWL) of 100.60 m at the pump well is assumed. The channel bottom elevation of 99.50 m is determined to provide channel flow velocities in a range of 0.2–1.3 m/s for the flow range between the minimum and maximum day flow rates. The head requirements for the sample bar screen system is in the range of 0.17–0.36 m (0.56–1.2 ft) when the pump wet well level is at the maximum elevation of 100.60. 22.4.2.2 Grit tanks. Process criteria. Grit, consisting of sand, gravel, cinders, and other heavy solid materials, is present in wastewater conveyed by either separate or combined sewer systems, with far more in the latter. Grit removal prevents unnecessary abrasion and wear of mechanical equipment, grit deposition in pipelines and channels, and accumulation of grit in primary sedimentation basins or aeration basins and anaerobic digesters. Traditionally removal of 95 percent of grit particles larger than 0.21 mm (0.008 in or 65 mesh) has been the target of grit equipment design. Modern designs are now capable of removing up to 75 percent of 0.15 mm (0.006 in or 100 mesh) to avoid adverse effects on downstream processes. A variety of grit removal devices have been applied over the years. The basic types of grit removal processes include aerated grit chambers, vortex-type, detritus tank, horizontal flow type and hydroclone. Vortex systems are increasingly being selected. Detritus



Chapter Twenty-Two

FIGURE 22.22 Schematic diagram of bar screen system.

22.44




tanks and aerated grit chambers are still popular. Depending on the type of grit removal process used, the removed grit is often further concentrated in a cyclone, classified, and then washed to remove light organic material captured with the grit. Key hydraulic design parameters. The key hydraulic design parameters for grit tanks include the inlet channel or inlet baffle, and effluent weir. Inlet channel/inlet baffle. For aerated grit chambers, the tank inlet and outlet should be positioned so that the flow through the tank is perpendicular to the roll pattern created by the diffused air. Inlet and outlet baffles serve to dissipate energy and minimize short circuiting. For vortex tanks, the flow into the vortex tank should be straight, smooth and streamlined. As a good practice, the straight inlet channel length should be seven times the width of the inlet channel or 15 ft, whichever is greater. The ideal velocity in the influent channel ranges from 0.6 to 0.9 m/s (2–3 ft/s) and should be used for flows between 40 and 80 percent of the peak flow. The minimum acceptable velocity for low flow is 0.15 m/s (0.5 ft/s). A baffle, located at the entrance, helps control the flow system in the tank and also forces the grit downward as it enters the tank. For detritus tanks, the performance relies on well-distributed flow into the settling basin. Allowances for inlet and outlet turbulence, as well as short circuiting, are necessary to determine the total tank area required. For horizontal flow grit chambers, velocity control throughout the chamber at approximately 0.3 m/s (1 ft/s) is important. An allowance for inlet and outlet turbulence is necessary to determine the actual length of the channel.

TABLE 22.8

Example Hydraulic Calculation of a Typical Bar Screen System Initial Operation Parameter

Min Day

1. Wastewater flow rate, Q (m3/s) 1.0 (mgd) 23 Bar screens Total number of units 3 Number of units in operation 2 Number of units on standby 1 Flow rate per screen in operation, q (m3/s) 0.5 Width of each bar screen, w (m) 2.5 2. At point 8 Pump wetwell HGL at high water level, HGL7 (m) 100.60 (pump starts at EL 100.60 and stops at EL 100.00) Pump well bottom EL (m) 99.00 Critical depth in a rectangular channel, Yc=(q2/g/w2)1/3 0.16 Bar screen channel depth= 1.10 pump WW HGL - channel bottom EL (m) (Water level at pump well controls upstream hydraulics if bar screen channel depth is higher than Yc) Is bar screen channel depth higher than Yc? yes 3. Point 8 to point 7 Channel bottom EL (m) 99.50

Design Operation

Avg.Day

Avg.Day

Max Hour Max Hour

1.6 36

2.0 46

3.2 73

3.2 73

3 2 1 0.8 2.5

3 2 1 1.0 2.5

3 2 1 1.1 2.5

3 2 1 1.6 2.5

100.60

100.60

100.60

100.60

99.00

99.00

99.00

99.00

0.22 1.10

0.25 1.10

0.26 1.10

0.35 1.10

yes

yes

yes

yes

99.50

99.50

99.50

99.50



22.46

Chapter Twenty-Two

TABLE 22.8


Min Day

Depth in channel, y7 (m) 1.10 Velocity, V7 (m/s) 0.18 Exit loss from channel to pump well Exit loss coeficient, Kexit 1.0 Headloss, Hle7=K Kexit. V772/2g (m) 0.00 HGL at point 7, HGL7 HGL8+Hle7 (m) 100.60 4. Point 7 to Point 6 Friction headloss through channel Length of approach channel, L6 (m) 7 Manning’s number for concrete channel, n 0.013 Channel width, w6 (m) 2.50 Water depth, h6 (m) 1.10 Velocity, V6 (m/s) 0.18 Hydraulic radius, R6 (h6 w6)/(2 h6 w6) 0.59 Headloss, Hlf6 (V6 n/r662/3)2 L6 (m) 0.00 HGL at Point 6, HGL6 HGL7 + Hlf6 (m) 100.60 5. Point 6 to Point 5 Bar width (m) 0.010 Bar shape factor, bsf 2.42 Cross-sectional width of bars, w (m) 0.89 Clear spacing of bars, b (m) 1.61 Upstream velocity head, h (m) 0.0041 Angle of bar screen with horizontal, p (degrees) 60 (Kirschmer’s eq),. Hls bsf w/b 1.33 h sin p (m) 0.01 Allow 0.15 m head for blinding by screenings, Ha (m) 0.15 HGL upstream of bar screen, HGL5 HGL6 Hls Ha (m)

100.76

6. Point 5 to Point 4 Friction headloss through channel Length of approach channel, L4 (m) 7.00 Manning’s number for concrete channel n 0.013 Channel width, w4 (m) 2.50 Channel bottom elevation (m) 99.65 Water depth, h4 (m) 1.11 Channel velocity, V V4 (m/s) 0.18 Hydraulic radius R4 h4 w4/(2 h4 w4) 0.59 Headloss , Hlf4 f (V4* V n/R / 4 (2/3) 2 L4 (m) 0.00 HGL at Point 4, HGL4 HGL5 + Hlf4 f (m) 7. Point 4 to Point 3 Headloss at sluice gate contraction Kgate Sluice gate width (m) Sluice gate height (m) Velocity through sluice gate, Vs (m/s)

Avg.Day

Design Operation Avg.Day

Max Hour Max Hour

1.10 0.29

1.10 0.36

1.10 0.39

1.10 0.58

1.0 0.00 100.60

1.0 0.01 100.61

1.0 0.01 100.61

1.0 0.02 100.62

7 0.013 2.50 1.10 0.29 0.59 0.00 100.60

7 0.013 2.50 1.11 0.36 0.59 0.00 100.61

7 0.013 2.50 1.11 0.39 0.59 0.00 100.61

7 0.013 2.50 1.12 0.57 0.59 0.00 100.62

0.010 2.42 0.89 1.61 0.0104 60

0.010 2.42 0.89 1.61 0.0163 60

0.010 2.42 0.89 1.61 0.0186 60

0.010 2.42 0.89 1.61 0.0418 60

0.02

0.03

0.03

0.06

0.15

0.15

0.15

0.15

100.77

100.78

100.79

100.83

7.00 0.013 2.50 99.65 1.12 0.29

7.00 0.013 2.50 99.65 1.13 0.35

7.00 0.013 2.50 99.65 1.14 0.38

7.00 0.013 2.50 99.65 1.18 0.54

0.59

0.59

0.60

0.61

0.00

0.00

0.00

0.00

100.76

100.77

100.78

100.79

100.83

1.0 1.2 0.9 0.38

1.0 1.2 0.9 0.59

1.0 1.2 0.9 0.74

1.0 1.2 0.9 0.78

1.0 1.2 0.9 1.13





Sluice gate headloss, Hls Kgate Vs2 /2g (m) HGL at Point 3, HGL3 (m) 8. Point 3 to Point 2 Water depth at point 2, h2 (m) Channel width, w2 (m) Channel velocity, V V2 (m/s) Fitting headloss through 45° bend Kbend 0.2 Headloss, Hlb2 = Kbend V 2 2/2g (m) Friction headloss through channel Length of approach channel, L2 (m) Manning’s, number for concrete channel n Hydraulic radius R2 h2 w2/(2 f2 f w2) (m) Headloss Hlf2 f (V2 V n/R / 2(2/3) 2 L2, (m) Entrance loss Kent 0.5 n Headloss, Hle2 = Kent V 2 2/2g(m) HGL at Point 2, HGL2 HGL3 Hlb2 Hlf2 f Hle2 (m) 9. Point 2 to Point 1 HGL at point 1, HGL 1 HGL2 (m) Invert EL of inlet sewer, INV1 (m) Crown EL of inlet sewer, CWN1 (m) Surcharge to inlet sewer?

Min Day

Avg.Day

Design Operation Avg.Day

Max Hour Max Hour

0.01

0.02

0.03

0.03

0.06

100.77

100.79

100.81

100.82

100.90

1.12 2.00 0.22 0.20 0.0005

1.14 2.00 0.35 0.20 0.0013

1.16 2.00 0.43 0.20 0.0019

1.17 2.00 0.46 0.20 0.0021

1.25 2.00 0.64 0.20 0.0042

4.00 0.013

4.00 0.013

4.00 0.013

4.00 0.013

4.00 0.013

0.53

0.53

0.54

0.54

0.56

0.0001

0.0002

0.0003

0.0003

0.0006

0.50 0.0013

0.50 0.0031

0.50 0.0047

0.50 0.0053

0.50 0.0105

100.77

100.79

100.82

100.83

100.91

100.77 99.50 101.65 No

100.79 99.50 101.65 No

100.82 99.50 101.65 No

100.83 99.50 101.65 No

100.91 99.50 101.65 No

Effluent weir. The effluent weir of the grit chamber provides the hydraulic control point of this process. With a free fall at the weir, critical depth occurs upstream near the weir and it affects the water surface profile upstream if the flow is subcritical. The effluent weir should be designed to keep the velocity below 0.3 m/s (1 ft/s) and to minimize turbulence in the outlet. Hydraulic design example. A schematic diagram of a typical vortex grit tank system is shown in Fig. 22.23. The effluent from the bar screen is pumped to the grit tank influent channel. The influent is distributed to three grit tanks. The hydraulic loading conditions are the same as those for the bar screens. Design hydraulic calculations for the vortex grit tank system is shown in Table 22.9. The head requirements for the sample grit tank system are in the range of 0.30–0.69 m (1.0–2.3 ft). 22.4.2.3 Sedimentation tanks. Process criteria. A typical municipal wastewater treatment system consists of primary sedimentation and secondary (or final) sedimentation tanks. The purpose of both type of sedimentation tanks is to separate the settleable solids from the liquid stream by gravity settling.



22.48

Chapter Twenty-Two

The primary sedimentation tank receives the wastewater passed through bar screens and/or grit tanks. The objectives of primary sedimentation are to produce a liquid effluent suitable for downstream biological treatment and to achieve solids separation. The solids result in a sludge that can be conveniently and economically treated before ultimate disposal. On an average basis, the primary sedimentation tank removes approximately 60 and 30 percent of influent total suspended solids (TSS) and 5-day biological oxygen demand (BOD5), respectively. The secondary sedimentation tank receives mixed liquor from the aeration tank. Mixed liquor is a suspended biological growth stream containing microorganisms and treated wastewater. The microorganisms settle with other settleable solids and the clear water is discharged from the sedimentation tank as an effluent. The sedimentation process also thickens the settled solids, a major part of which is returned to the aeration tank and the remainder is wasted as secondary sludge. Sedimentation tank performance is critical for meeting effluent limits for TSS and BOD5. The secondary sedimentation effluents are usually designed to produce 30 mg/L or lower for TSS or BOD5, depending on the effluent requirement. Both primary and secondary sedimentation tanks are commonly arranged in either rectangular or circular shape. Key design parameters include surface overflow rate (SOR), tank water depth, hydraulic detention time, and weir loading rate. Solids loading rate is another important parameter for the secondary sedimentation tank. A properly designed sedimentation tank will provide similar performance for both rectangular and circular shapes. Choice of the shape depends on the site constraints, construction cost, and designer preference. Key hydraulic design parameters. The key hydraulic design parameters for sedimentation tanks include the inlet conditions, inlet channel, inlet flow distribution, inlet baffle, outlet conditions, overflow weir, and effluent launder. Inlet conditions. Inlets should be designed to dissipate the inlet port velocity, distribute flow and solids equally across the cross-sectional area of the tank, and prevent short circuiting in the sedimentation tank. The minimum distance between the inlet and outlet should be 3 m (10 ft) unless the tank includes special provisions to prevent short circuiting. Inlet channel. Inlet channels should be designed to maintain velocities high enough to prevent solids deposition. The minimum channel velocity is typically 0.3 m/s (1 ft/s). Alternatively, inlet channel aeration or water jet nozzles can be designed to prevent solids deposition. Inlet flow distribution. Inlet flow can be distributed by inlet weirs, submerged ports, or orifices with velocities between 0.05 and 0.15 m/s (0.15–0.5 ft/s), and sluice gates or gate valves. Uniform flow to the sedimentation tanks can be achieved by locating inlet ports away from sides, adding partitions or baffles in the inlet zone to redirect the influent, and creating a higher head loss in the inlet ports relative to that in the inlet channel. Alternatively, splitter boxes are used for equally splitting the flow as well as solids contained in the liquid into multiple sedimentation tanks. Inlet baffle. Inlet baffles are designed to dissipate the energy of the inlet velocities. Baffles are usually installed 0.6–0.9 m (2–3 ft) downstream of the inlet port and submerged 0.45–0.6 (1.5–2 ft), depending on tank depth. The top of the baffle should be far enough below the water surface to allow scum to pass over the top. Circular tanks typically have a feed well with a diameter 15 to 20 percent of the tank diameter. The submergence varies depending on the manufacturer. Outlet conditions. Effluent should be uniformly withdrawn to prevent localized high velocity zones and short circuiting. Typically, effluent is withdrawn from a sedimentation




FIGURE 22.23 Schematic diagram of vortex grit tank system.

TABLE 22.9

Example Hydraulic Calculation of a Typical Vortex Grit Tank System Initial Operation Parameter

1. Wastewater flow rate, Q (m3/s) (mgd) 2. Vortex grit tanks Total number of units Number of units in operation Number of units on standby Flow rate per vortex grit tank in operation (m3/s)

Design Operation

Min Day

Avg.Day

Avg.Day

Max Hour

Peak

1.0 23

1.6 36

2.0 46

3.2 73

3.2 73

3 2 1

3 2 1

3 2 1

3 3 0

3 2 1

0.5

0.8

1.0

1.1

1.6

106.00 105.00 0.8

106.00 105.00 1.0

106.00 105.00 1.1

106.00 105.00 1.6

Control point is located at Point 8 (effluent channel weir) Hydraulic calculations upstream of control point 3. At Point 8 Headloss over sharp-crested weir Sharp-crested weir EL, weir EL (m) Effluent channel bottom EL (m) Flow rate over weir, q (m3/s)

106.00 105.00 0.5



22.50

Chapter Twenty-Two

TABLE 22.9


Min Day

Design Operation

Avg.Day

Avg.Day

Max Hour

Peak

3.00

3.00

3.00

3.00

3.00

0.20 0.20 0.00

0.28 0.28 0.00

0.32 0.32 0.00

0.34 0.34 0.00

0.45 0.45 0.00

106.20

106.28

106.32

106.34

106.45

3.00 105.00 1.20

3.00 105.00 1.28

3.00 105.00 1.32

3.00 105.00 1.34

3.00 105.00 1.45

1.0 0.0010

1.0 0.0022

1.0 0.0032

1.0 0.0036

1.0 0.0069

HGL at Point 7, HGL7 HGL8 Hle7 (m) 106.20

106.28

106.33

106.34

106.45

2.50 105.00 1.28 0.25

2.50 105.00 1.33 0.30

2.50 105.00 1.34 0.32

2.50 105.00 1.45 0.44

10.00 0.013

10.00 0.013

10.00 0.013

10.00 0.013

0.63 0.0002

0.64 0.0003

0.65 0.0003

0.67 0.0006

1.0 0.0032

1.0 0.0046

1.0 0.0051

1.0 0.0099

106.28

106.33

106.35

106.46

1.0 1.5 1.0 1.20

1..0 1.5 1.0 1.28

1.0 1.5 1.0 1.33

1.0 1.5 1.0 1.34

1.0 1.5 1.0 1.45

1.0

1.0

1.0

1.0

1.0

0.33

0.53

0.67

0.71

1.07

Length of weir, L (m) Head over end contracted weir, He (assumed) Headloss, He8 (q/1.84 (L – 0.2He)(2/3) (m) Hle8 – He (must be zero) HGL at Point 8, HGL8 weir EL Hle8 (m) 4. Point 8 to Point 7 Channel width, w7 (m) Channel bottom EL (m) Water depth, h7 (m) Velocity, V 7 (m/s) Exit headloss from channel to effluent weir Exit headloss coefficient Kexit 1.0 Headloss, Hle7 Kexit V72/2g (m)

5. Point 7 to Point 6 Channel width, w6 (m) 2.50 Channel bottom EL (m) 105.00 Water depth, h6 (m) 1.20 Velocity, V V6 (m/s) 0.17 Friction headloss through channel Length of approach channel, L6 (m) 10.00 Manning’s number for concrete channel n 0.013 Hydraulic radius, R6 (h6 w6)/ (2 x h6 w6) (m) 0.61 2 Headloss Hlf6 f [(V6 V n/R / 6 (2/3)] L6(m)0.0001 Fitting headloss through 90º bend Fitting headloss coefficient Kbend 1.0 1.0 Headloss, Hlb6 Kbend V6 V 2/2g(m) 0.0014 HGL at Point 6, HGL6 HGL7 Hlf6 f Hlb6 (m) 106.21 6. Point 6 to Point 5 Headloss through sluice gate Sluice gate headloss coefficient Kgate 1.0 Sluice gate width (m) Sluice gate height (m) Water depth, h5 (m) Sluice gate height or h5, whichever is smaller (m) Velocity through sluice gate, V5 (m/s) V





Headloss, Hls5 Kgate V5 V 2/2g (m) HGL at Point 5, HGL5 HGL6 Hls5 (m)

Avg.Day

Avg.Day

Max Hour

0.0057

0.0145

0.0227

0.0258

0.0580

106.21

106.30

106.36

106.37

106.52

2.50 105.20 1.01 0.29

2.50 105.20 1.16 0.35

2.50 105.20 1.17 0.36

2.50 105.20 1.32 0.48

1.0 0.0043

1.0 0.0061

1.0 0.0067

1.0 0.0120

10.00 0.013

10.00 0.013

10.00 0.013

10.00 0.013

0.58 0.0003

0.60 0.0004

0.61 0.0004

0.64 0.0007

106.30

106.36

106.38

106.54

0.06

0.06

0.06

0.06

0.06

106.27

106.36

106.42

106.44

106.60

2.00 105.60 0.67 0.37

2.00 105.60 0.76 0.52

2.00 105.60 0.82 0.61

2.00 105.60 0.84 0.63

2.00 105.60 1.00 0.80

14.00 0.013

14.00 0.013

14.00 0.013

14.00 0.013

14.00 0.013

0.40

0.43

0.45

0.46

0.50

0.0011

0.0020

0.0025

0.0027

0.0039

1.0 1.5 1.0

1..0 1.5 1.0

1.0 1.5 1.0

1.0 1.5 1.0

1.0 1.5 1.0

7. Point 5 to Point 4 Channel width, w4 (m) 2.50 Bottom of channel EL (m) 105.20 Water depth, h4 (m) 1.01 Channel velocity, V V4 (m/s) 0.20 Fitting headloss through a 90º bend Fitting headloss coefficient Kbend 1.0 1.0 Headloss, Hlb4 Kbend V4 V 2/2g (m) 0.0020 Friction headloss through channel Length of channel, L4 (m) 10.00 Manning’s n for concrete channel 0.013 Hydraulic radius, R4 h4 w4/ (2 h4 w4) (m) 0.56 Headloss, Hlf4 f [(V4 V n/R / 4(2/3)]2 L4 (m)0.0001 HGL at Point 4, HGL4 HGL5 Hlb4 Hlf4 f (m) 106.21 8. Point 4 to Point 3 Headloss across vortex grit tank, Hltank (m) (per manufacturer recommendations) HGL at Point 3, HGL3 HGL4 Hltank (m) 9. Point 3 to Point 2 Channel width, w2 (m) Bottom of channel EL (m) Water depth, h2 (m) Channel velocity, V V2 (m/s) Friction headloss through channel Length of approach channel, L2 (m) Manning’s n for concrete channel Hydraulic radius, R2 h2 w2/ (2 h2 w2) (m) 2 Headloss, Hlf2 f [(V2 V n/R / 2(2/3)] L2 (m) Headloss through sluice gate Sluice gate headloss coefficient Kgate 1.0 Sluice gate width (m) Sluice gate height (m)

Design Operation

Min Day

Peak



22.52

Chapter Twenty-Two

TABLE 22.9

(Continued) Initial Operation

Design Operation

Parameter

Min Day

Avg.Day

Avg.Day

Max Hour

Peak

Water depth, h2 (m) Sluice gate height or h2, whichever is smaller Velocity through sluice gate,V V2 (m/s) Headloss, Hls2 Kgate V2 V 2/2g (m) HGL at Point 2, HGL2 HGL3 Hlf2 f Hls2 (m)

0.67

0.76

0.82

0.84

1.00

0.67 0.49 0.0125

0.76 0.70 0.0249

0.82 0.81 0.0335

0.84 0.85 0.0364

1.00 1.07 0.0586

106.29

106.39

106.46

106.48

106.66

2.00 105.65 0.64 0.39

2.00 105.65 0.74 0.54

2.00 105.65 0.81 0.62

2.00 105.65 0.83 0.64

2.00 105.65 1.01 0.79

1.0

1.0

1.0

1.0

1.0

0.0078

0.0149

0.0195

0.0210

0.0322

5.00 0.013

5.00 0.013

5.00 0.013

5.00 0.013

5.00 0.013

10. Point 2 to Point 1 Channel width, w1 (m) Bottom of channel EL (m) Water depth, h1 (m) Channel velocity, V1 (m/s) Fitting headloss through a 90º bend Fitting headloss coefficient Kbend 1.0 Headloss, Hlb1 Kbend 2 V1 /2g (m) Friction headloss through channel Length of approach channel, L1 (m) Manning’s n for concrete channel Hydraulic radius, R1 h1 w1/ (2 h1 w1) (m) (2/3)2 Headloss, Hlf1 f (V1 n/R / 1 L1 (m)

0.39

0.43

0.45

0.45

0.50

0.0005

0.0008

0.0009

0.0010

0.0013

(Influent channel may be aerated using diffused air to prevent solids settling or odor problem) HGL at Point 1, HGL1 HGL2 Hlc1 Hlf1 f (m)

106.30

106.41

106.48

106.50

106.69

tank over an effluent weir into a trough and/or effluent channel. Clarifier performance can often be improved by installation of interior baffles. For circular tanks, particularly for secondary sedimentation tanks, a baffle mounted on the wall beneath the effluent weir can deflect solids rising along the wall. Alternatively, mid-radius baffles supported by the sludge removal mechanism are also available. Overflow weir. The overflow weir must be level to promote uniform effluent withdrawal. Weirs may be either straight edged or “V”-notched. “V”-notched weirs have higher headloss, but provide better lateral distribution than straight-edged weirs that are imperfectly leveled. Effluent launder (or trough). Effluent launders may be designed with submerged orifices or free discharge into the collection chamber or channel from which the effluent flows to the effluent pipe. Disadvantage of the submerged launder is that it is not effective




in varying flow rates. Disadvantage of the free fall launder is potential release of odorous gases. Two principal approaches to weir and launder design are the long-launder and short-launder options. Long launders control the head loss over the weir within a narrow range. In cold regions, fluctuating water levels with short launders would minimize ice attachment to launders and basin walls. Hydraulic design example for primary sedimentation. A schematic diagram of typical circular primary sedimentation tank system is shown in Fig. 22.24. The primary sedimentation tanks receive the grit tank effluent and hydraulic loading conditions are the same as those of the grit tanks. A single primary sedimentation tank is shown for simplicity. Design hydraulic calculations for the primary sedimentation tank system is shown in Table 22.10. Note that the design locates Points 5 and 6 at elevations such that downstream flow conditions will not impact flow conditions in the effluent channel or overflow weir. The head requirements for the sample primary sedimentation tanks are in the range of 1.1–1.5 m (3.6–4.9 ft). Hydraulic design example for secondary sedimentation. A schematic diagram of typical rectangular secondary sedimentation tank system is shown in Figure 22.25. The secondary sedimentation tanks receive flows from the aeration tanks and hydraulic loading conditions are same as those of the aeration tanks. A single secondary sedimentation tank is shown for simplicity. Design hydraulic calculations for the secondary sedimentation tank system is shown in Table 22.11. The head requirements for the sample secondary sedimentation tanks are in the range of 1.6–1.7 m (5.2–6.2 ft). 22.4.2.4 Aeration tanks Process criteria. The most common aerobic suspended growth treatment system for municipal wastewater is the activated sludge system. Wastewater and biological solids (mixed–liquor suspended solids or MLSS) are combined, mixed, and aerated in the aeration tank. The biological MLSS solids take up the organics and nutrients contained in the wastewater and convert them into more biosolids and gaseous by-products. After sufficient time for biological reactions, the mixed liquor is transferred to the following secondary sedimentation tanks where biosolids are separated from the wastewater. The separated wastewater is discharged as an effluent. The separated biosolids are returned to the aeration tank (return activated sludge or RAS) while a predetermined amount of the separated biosolids is wasted as waste activated sludge (WAS). Factors that must be considered in the design of the activated sludge process include loading criteria, selection of reactor type, sludge production, oxygen requirements and transfer, nutrient requirements, environmental requirements, solid-liquid separation, and effluent characteristics. Sizing of aeration basins is based on two key factors: providing sufficient time for oxidation of organics or ammonia nitrogen; and maintaining of a flocculent, well-settling MLSS that can be effectively removed by gravity settling. Solids residence time (SRT) or mean cell residence time (MCRT) is often used to relate substrate removal time requirements to biological growth and biosolids production. Once an SRT is selected, calculation of aeration tank volume requires an estimation of biosolids production and selection of proper MLSS concentration. The selected MLSS concentration along with the solids settling characteristics is important to the final sedimentation tank performance. Therefore, sizing of the aeration tank is always optimized with the final sedimentation tank design. The aeration tank should be provided with sufficient oxygen required for the biological reaction and sufficient power required for thorough mixing of the biomass with the incoming wastewater stream. Although a variety of diffused aeration and mechanical aeration systems are available, diffused aeration systems are more popular in the municipal wastewater treatment. Downloaded from Digital Engineering Library @ McGraw-Hill (www.digitalengineeringlibrary.com) Copyright © 2004 The McGraw-Hill Companies. All rights reserved. Any use is subject to the Terms of Use as given at the website.


Chapter Twenty-Two

FIGURE 22.24 Schematic diagram of primary sedimentation tank (PST) system.

22.54




Aeration basin configurations.. Common aeration basins include various process configurations, physical configurations and designs for process selectors. A schematic diagram of a typical rectangular aeration tank system is shown in Fig. 22.26. Process configuration Various aeration process configurations can be used depending on the range of loading conditions, design effluent quality, aeration system design requirements and flexibility of operation. Configurations often encountered include complete mix, plug flow, oxidation ditch, and a combination of these. For smaller plants, oxidation

TABLE 22.10 Tank System

Example Hydraulic Calculations of a Typical Primary Sedimentation Initial Operation Parameter

1. Wastewater flow rate, Q (m3/s) (mgd) 2. Primary sedimentation tanks (PSTs) Total number of units Number of units in operation Number of units on standby Flow rate per PTS in operation, q (m3/s)

Min Day Avg.Day

Design Operation Avg.Day Max Hour

Peak

1.0 23

1.6 36

2.0 46

3.20 73

3.20 73

3 2 1

3 2 1

3 3 0

3 3 0

3 2 1

0.5

0.8

0.7

1.1

1.6

104.46

104.46

104.46

104.46

104.46

0.10

0.10

0.10

0.10

0.10

104.56

104.56

104.56

104.56

104.56

45.0 2 0.50 0.25

45.0 2 0.80 0.40

45.0 2 0.67 0.33

45.0 2 1.07 0.53

45.0 2 1.60 0.80

0.20 1.00

0.20 1.00

0.20 1.00

0.20 1.00

0.20 1.00

69.87

69.87

69.87

69.87

69.87

0.14

0.14

0.14

0.14

0.14

Control points are located at Points 5 and 6 so that back up from down stream does not flood effluent channel or overflow weir. Hydraulic Calculations beginning at Point 7 1. At Point 7 HGL7 must be equal to HGL1 of aeration tank (m) 2. At Point 6 Allowance of 0.10 m from HGL at pipe entrance to bottom of PST effluent trough at discharge end (m) Elevation of PTS trough bottom at discharge end, ELdcb (m) Calculation of water depth in PST effluent trough Tank diameter, Dt (m) Number of channels per tank, nc Total flow through tank, q (m3/s) Flow per channel, qc q/nc (m3/s) Channel slope, Sc, (selected to prevent solids settling) Channel width, w6 (m) Channel length, Lc 3.14 (Dt (w6/2))/nc (m) Change in channel EL, EL dif Sc Lc (m)



22.56

Chapter Twenty-Two

TABLE 22.10


Min Day Avg.Day

Critical depth, yc (qc2/(g w62)0.33 (m) 0.19 Water depth at upstream end of channel, yu 2 (yc)2 (yc (S*L/3) L 2]0.5 (2 Sc L L/3) (m) 0.21 Channel bottom El at upstream end of trough, 104.70 ELucb ELdcb ELdif (m) HGL at trough downstream, HGL6d ELdcb yc (m) HGL at trough upstream, HGL6u ELucb yu (m) 3. Point 6 to Point 5 Allowance to Weir from high trough HGL (m) Weir elevation, Elwe, max. HGL6u allowance (m) Headloss over V V–notch weirs Number of weirs per tank, Nw Tank diameter, Dt, (m) Weir length, Lw (Dt) 3.14 (m) Hydraulic load, So q/Lw / , [(m3·/s)/m] Weir angle, A, (degrees) V-notch height, Vh (m) V-notch width, Vw 2 (TAN(A ( /2) Vh (m) Space between notches, Esv (m) Number of notches per weir, nv Lw/(Ew Esv) Flow per notch, Qcw q/nv (m3/s) Weir coefficient for 90º notch, Cw Water depth over the weir, hle5 (Qcw/Cw)(1/2.48) hle5 < Vh? (If not, need to readjust calculations) HGL at Point 5, HGL5 ELwe hle5 (m) 4. Point 5 to Point 4 Headloss through primary sedimentation tanks Number of tanks, Nt Flow per tank, q (m3/s) Tank diameter, Dt (m) Side water depth, Dsw (m) Tank bottom elevation, ELt HGL5 Dsw (m) Tank floor slope, St (%)


Peak

0.26

0.23

0.31

0.41

0.33

0.28

0.42

0.58

104.70

104.70

104.70

104.70

104.75

104.82

104.79

104.87

104.97

104.91

105.03

104.98

105.12

105.28

0.10

0.10

0.10

.010

0.10

105.38

105.38

105.38

105.38

105.38

1 45.00 141.30 0.0035 90.00 0.10

1 45.00 141.30 0.0057 90.00 0.10

1 45.00 141.30 0.0047 90.00 0.10

1 45.00 141.30 0.0075 90.00 0.10

1 45.00 141.30 0.0113 90.00 0.10

0.20 0.03

0.20 0.03

0.20 0.03

0.20 0.03

0.20 0.03

614 0.0008 1.34

614 0.0013 1.34

614 0.0011 1.34

614 0.0017 1.34

614 0.0026 1.34

0.05

0.06

0.06

0.07

0.08

Yes

Yes

Yes

Yes

Yes

105.44

105.45

105.44

105.45

105.47

2 0.50 45.00 4.30

2 0.80 45.00 4.30

3 0.67 45.00 4.30

3 1.07 45.00 4.30

2 1.60 45.00 4.30

101.14 8.33

101.14 8.33

101.14 8.33

101.14 8.33

101.14 8.33





Minimum floor tank elevation, ELtf 0.0833 (Dt/2) t EL (m) Headloss through tank, hlt4 t (m) (available from equipment manufacturer) HGL at Point 4, HGL4 HGL5 hlt4 t (m)

Min Day Avg.Day


Peak

99.27

99.27

99.27

99.27

99.27

0.05

0.05

0.05

0.05

0.05

105.49

105.50

105.49

105.50

105.52

1.07 6.50

1.07 6.50

1.07 6.50

1.07 6.50

0.89 120 0.27

0.74 120 0.27

1.19 120 0.27

1.78 120 0.27

5. Point 4 to Point 3 Headloss through PST influent pier Pier diameter, Dp 1.07 m 1.07 Pier length, Lp (m) 6.50 Velocity, V3 V Q/(3.14 (Dp/2)2) (m/s) 0.56 Hazen-Williams coefficient, Cp 120 Hydraulic radius, Rp Dp/4 (m) 0.27 Slope, Sp [V3/(0.85 V Cp Rp(0.63)](1/0.54) (%) 0.03 Headloss, Hlf3 f Lp Sp (m) 0.0020 Exit headloss from pier Exit headloss coefficient Kexit 1.0 1 Headloss, hle3 K V3 V 2/2g (m) 0.0158

0.07 0.0047

0.05 0.0033

0.12 0.0079

0.26 0.0168

1 0.0404

1 0.0281

1 0.0719

1 0.1617

HGL at Point 3, HGL3 HGL4 Hlf3 f hle3 (m)

105.54

105.52

105.58

105.69

3

3

3

3

1 1.20 0.80 0.71

1 1.20 0.67 0.59

1 1.20 1.07 0.94

1 1.20 1.60 1.42

120 0.30 70.0

120 0.30 70.0

120 0.30 70.0

120 0.30 70.0

0.04 0.0287

0.03 0.0205

0.07 0.0490

0.15 0.1037

0.05 0.0128

0.05 0.0089

0.05 0.0227

0.05 0.0511

105.58

105.55

105.65

105.85

105.50

6. Point 3 to Point 2 Total number of pipes 3 Number of pipes per primary sedimentation tank 1 Pipe diameter, Dp (m) 1.20 Flow per pipe, q (m3/s) 0.50 Velocity, V V2 0.44 Friction headloss through primary sedimentation tank influent pipe Hazen-Williams coefficient, Cp 120 Hydraulic radius, Rp Dp/4 (m) 0.30 Length of pipe, Lp (m) 70.0 Slope, Sp [V2/(0.85 V Cp Rp(0.63)](1/0.54) (%) 0.02 Headloss, hlf2 f Lp Sp (m) 0.0120 Fitting headloss through two 45º bends Fitting headloss coefficient Kbend 0.5 0.05 Headloss, hlb2 K V2 V 2/2g (m) 0.0050 HGL at Point 2, HGL2 HGL3 hlb2 hlf2 f (m)

105.52

7. At Point 1



22.58

Chapter Twenty-Two

TABLE 22.10


Entrance headloss from primary sedimentation tank influent distribution box to influent pipe Pipe diameter, Dp (m) Flow per pipe, q (m3/s) Velocity, V1 (m/s) Entrance headloss coefficient Kentrance 0.5 Headloss, Hle1 Kentrance V12/2g (m) HGL at Point 1, HGL1 HGL2 Hle1 (m) Allowance to grit tank effluent weir from maximum HGL1, Hall (m) Grit tank effluent elevation, ELgr HGL1 Hall (m)

Min Day Avg.Day


Peak

1.20 0.50 0.44

1.20 0.80 0.71

1.20 0.67 0.59

1.20 1.07 0.94

1.20 1.60 1.42

0.50

0.50

0.50

0.50

0.50

0.0050

0.0128

0.0089

0.0227

0.0511

105.52

105.60

105.56

105.68

105.90

0.10

0.10

0.10

0.10

0.10

106.00

106.00

106.00

106.00

106.00

ditches are more popular and for larger plants, plug flow is favored. Various modifications of plug flow systems include conventional, tapered aeration, step aeration, modified aeration, and contact stabilization. Physical configuration. Various physical configurations are used in the aeration tank design, including rectangular, circular, oval, and octagonal shapes. Selector design. Selectors are small compartments for aerobic, anoxic or anaerobic processing usually located in the front end of the aeration tank. The purpose of the selectors is to promote the growth of floc-forming microorganisms by providing a favorable food to microorganisms (F:M) ratio while suppressing filamentous growth. Typically selectors are designed with low HRTs and high F:M ratio. Key hydraulic design parameters. The key hydraulic design parameters for aeration tanks include the distribution box, inlet channel, inlet flow distribution, inlet baffles, aeration equipment, RAS, effluent weir, and effluent channel. Distribution box. Sluice gates, weirs, gate valves or orifices installed in a distribution box are often used to distribute the upstream flow to multiple aeration tanks and to a secondary treatment bypass line. Design should provide the desired rate of flow distribution at all flow conditions with minimum headloss. Provisions to minimize solids deposition in the distribution box and appurtenances should be considered. Inlet channel. Inlet channels should be designed to maintain velocities high enough to prevent solids deposition but low enough to minimize headloss. A velocity of 0.3 m/s (1 ft/s) is typically used to keep organic solids in suspension. Alternatively, inlet channel aeration with diffused air, fed at a rate of 0.5–0.8 m3/min (20–30 scfm), is often used. Inlet flow distribution. Inlet flow can be distributed by inlet weirs, submerged ports or orifices, and sluice gates or gate valves. Return activated sludge may be introduced prior Downloaded from Digital Engineering Library @ McGraw-Hill (www.digitalengineeringlibrary.com) Copyright © 2004 The McGraw-Hill Companies. All rights reserved. Any use is subject to the Terms of Use as given at the website.


FIGURE 22.25 Schematic diagram of final sedimentation tank.




22.60

Chapter Twenty-Two

TABLE 22.11

Example Hydraulic Calculation of a Typical Final Sedimentation Tank System Initial Operation PARAMETER

1. Wastewater flow rate, Q (m3/s) (mgd) RAS flow, Qras (% of average day flow) RAS flow, Qras /100, (m3/s) Final sedimentation tank influent flow, Qin, (m3/s) Final sedimentation tank effluent flow, Qeff, f (m3/s) Final sedimentation tanks Total number of units Number of units in operation Number of units on standby Tank width (m) Influent per operating tank, qin, (m3/s) Effluent per operating tank, qeff, f (m3/s) 2. Select control point at Point 3 (where effluent wiers are located) Hydraulic calculations downstream of control point At Point 3 V-notch weir Number per tank, Nw Individual weir length, Lw (m) Total weir length, Lwt Lw Nw (m) Weir angle, A degrees V-notch height, Vh (m) V-notch width, Vw 2 (TAN( N(A/2) Vh (m) Space between notches, Esv (m) Total number of notches per tank, nv Lwt/( t Vw Esv) Flow per notch, Qcw qeff/ f nv Weir coefficient for 90º notch, Cw Water depth over the weir, hle3 (Qcw/Cw)(1/2.48) (m) hle3 Vh? (If not, need to readjust calculations) Weir EL (m) (Select weir elevation so that HGL1 equals aeration tank’s HGL6) EGL at Point 3, EGL3 Weir EL hle3 (m) Velocity head, HV3 0 (assume V3 V 0) (m)

Min Day Avg.Day


Peak

1.0 23

1.6 36

2.0 46

3.2 73

3.2 73

20

50

50

100

100

0.32

0.80

1.00

2.00

2.00

1.32

2.40

3.00

5.20

5.20

1.00

1.60

2.00

3.20

3.20

4 3 1 16

4 3 1 16

4 3 1 16

4 4 0 16

4 3 1 16

0.44

0.80

1.00

1.30

1.73

0.33

0.53

0.67

0.80

1.07

20 7.0 140.0 90.0 0.10

20 7.0 140.0 90.0 0.10

20 7.0 140.0 90.0 0.10

20 7.0 140.0 90.0 0.10

20 7.0 140.0 90.0 0.10

0.20 0.03

0.20 0.03

0.20 0.03

0.20 0.03

0.20 0.03

608 0.0005 1.34

608 0.0009 1.34

608 0.0011 1.34

608 0.0013 1.34

608 0.0018 1.34

0.04

0.05

0.06

0.06

0.07

yes

yes

yes

yes

yes

103.37

103.37

103.37

103.37

103.37

103.41

103.42

103.43

103.43

103.44

0.00

0.00

0.00

0.00

0.00




(Continued) Initial Operation PARAMETER

HGL at point 3, HGL3 Weir EL hle3 (m) 3. Point 3 to Point 4 Effluent troughs Number of troughs, nt Flow per trough, qt qeff/ f nt (m3/s) Trough slope, St (%) (select to prevent solids settling) Trough width, w6 (m) Approximate trough length, Lt (m) Change in trough EL due to slope difEL4 St* Lt (m) Critical depth at downstream end, yc (qt2/(gw62)0.33 (m) Water depth at upstream end of trough for free fall from trough into final effluent channel yu4 [2 (yc)2 (yc (S*L/3) L 2].5 (2 S L L/3) (m) Max water EL downstream of weir (occurring at max. hourly flow with one tank out of service) Elmax4 weir EL 0.1 (m) (see Point 3 for weirEL) Trough bottom EL at upstream end of trough, TbuEL4 (m) Tbu EL4 EL max4 yu for max hour flow with one tank out of service HGL at upstream end, HGL4u Tbu EL4 yu4 (m) Velocity head, HV4 V u0 (assume V 0) (m) EGL at upstream end, EGL4u HGL4u HV4 V u (m) Trough bottom EL at downstream end of trough Tbd EL4 Tbu EL4 dif EL4 (m) HGL at point 4, HGL4 TbdEL4 yc (m) Velocity head, HV4 V d Vc2/2g (m)

Min Day Avg.Day

Initial Operation

Avg.Day Max Hour

Peak

103.41

103.42

103.43

103.43

103.44

10 0.03

10 0.05

10 0.07

10 0.08

10 0.11

0.20 0.5 7.0

0.20 0.5 7.0

0.20 0.5 7.0

0.20 0.5 7.0

0.20 0.5 7.0

0.01

0.01

0.01

0.01

0.01

0.08

0.11

0.12

0.14

0.17

0.12

0.17

0.20

0.23

0.28

103.27

102.99

102.99

102.99

102.99

102.99

103.11

103.16

103.19

103.22

103.27

0.00

0.00

0.00

0.00

0.00

103.11

103.16

103.19

103.22

103.27

102.97

102.97

102.97

102.97

102.97

103.05 0.04

103.08 0.05

103.10 0.06

103.11 0.07

103.14 0.08



22.62

Chapter Twenty-Two

TABLE 22.11


EGL at downstream end, EGL4d HGL4d HV4 V d (m)

Min Day Avg.Day

Design Operation

Avg.Day Max Hour

Peak

103.09

103.13

103.16

103.18

103.22

102.87

102.87

102.87

102.87

102.87

HGL maximum at Point 5, HGL5ELmax5(m)102.87 Velocity head, HV5 V 0 (assume V 0) (m) 0.00 EGL maximum at Point 5, EGL5m HGL5m HV5 V (m) 102.87

102.87

102.87

102.87

102.87

0.00

0.00

0.00

0.00

102.87

102.87

102.87

102.87

1.00

1.60

2.00

3.20

3.20

0.20 3.0 64.0

0.20 3.0 64.0

0.20 3.0 64.0

0.20 3.0 64.0

0.20 3.0 64.0

0.13

0.13

0.13

0.13

0.13

0.23 0.29

0.31 0.43

0.36 0.52

0.49 0.74

0.49 0.74

102.13

102.13

102.13

102.13

102.13

102.42

102.56

102.65

102.87

102.87

0.00

0.00

0.00

0.00

0.00

4. Point 4 to Point 5 Effluent channel upstream Max. water surface level at upstream end of effluent channel, ELmax5 TbdEL4 0.1 (m)

5. Point 5 to Point 6 Effluent channel downstream Flow through channel, Qeff (m3/s) Channel slope, Sc (%) (select to prevent solids settling) Channel width, w6 (m) Approximate channel length, Lch (m) Change in channel EL, difEL6 Sc Lch (m) Critical depth, yc (q2/(gw62)0.33 (m) Water depth at upstream end of channel, yu6 [2 (yc)2 (yc (S L/3) L 2].5 (2 S L L/3) (m) Channel bottom EL at upstream end of channel, cbuEL6 HGL5- maximum yu6 (m) HGL at upstream end of channel, HGL5 cbuEL6 yu6 (m) Velocity head, HV5 V 0 (assume V 0) (m) EGL at upstream end of channel, EGL5 HGL5 HV5 V (m)

102.42

102.56

102.65

102.87

102.87

Channel bottom EL at downstream end of channel, cbdEL6 cbuEL6 difEL6 (m)

102.00

102.00

102.00

102.00

102.00

HGL at Point 6, HGL6 cbdEL6 yc (m) Velocity head, HV6 V Vc2/2g (m)

102.23 0.11

102.31 0.15

102.36 0.17

102.50 0.24

102.50 0.24





EGL at Point 6, EGL6 HGL6 HV6 V (m) 6. At Point 7 Max water EL downstream of channel end free-fall HGL at Point 7, HGL7 cbdEL6 0.1 (m) (This must be the same as maximum elevation at Point 1 of multimedia filter)

Min Day Avg.Day

Design Operation

Avg.Day Max Hour

Peak

102.34

102.47

102.54

102.74

102.74

101.90

101.90

101.90

101.90

101.90

3

3

3

4

3

0.44 16.0 120.0

0.80 16.0 120.0

1.00 16.0 120.0

1.30 16.0 120.0

1.73 16.0 120.0

99.2 4.24

99.2 4.25

99.2 4.26

99.2 4.26

99.2 4.27

0.0

0.0

0.0

0.0

0.0

103.41

103.42

103.43

103.43

103.44

0.00

0.00

0.00

0.00

0.00

103.41

103.42

103.43

103.43

103.44

1.0 1.0 1.0 4

1.0 1.0 1.0 4

1.0 1.0 1.0 4

1.0 1.0 1.0 4

1.0 1.0 1.0 4

0.11

0.20

0.25

0.33

0.43

0.21 1.0

0.31 1.0

0.36 0.9

0.44 0.9

0.53 0.9

0.16

0.24

0.27

0.33

0.40

0.17

0.31

0.38

0.49

0.66

0.11

0.20

0.25

0.33

0.44

Hydraulic Calculations Upstream of Control Point 7. At Point 2 Final sedimentation tanks (Gould type) Number of tanks in operation, nt Flow per tank upstream of sludge collection, qin (m3/s) Tank width, Wt (m) Tank length, Lt (m) Tank bottom elevation at influent end (m) Side water depth (m) Assume friction losses, Hlf2, f through tank are negligible EGL at Point 2, EGL2 EGL3 Hlf2 f (m) Velocity head, HV2 V 0 (assume V 0) (m) HGL at Point 2, HGL2 EGL3 HV2 V (m) 8. Point 2 to Point 1 Tank influent sluice gates Height (m) Width, Ws (m) Area (m2) Number of sluice gates per tank, Nsg Flow per sluice gate, qsg qin/Nsg / (m3/s) Upstream head over weir, Du (select so Qsub qsg D) (m) Effective sluice gate width, Ws' Ws (0.1)(2 contractions)(Dd) (m) Downstream head over weir, Dd (qsg/1.84/Ws')(2/3) (m) Free–fall flow, Qfree 1.84 Ws' Du(3/2), (m3/s) Submerged flow, Qsub Qfree (1 (Dd/ d/Du)3/2)0.385 (m3/s)



22.64

Chapter Twenty-Two

TABLE 22.11


Difference, (Qsub qsg) (m3/s) (should be zero) Head difference between tank and channel, Hl 1 Du Dd (m) Top of sluice gate set elevation, Els HGL2 Dd (m) HGL at Point 1 (upstream of sluice gate), HGL1 HGL2 Hl1 (m) Velocity head, HV1 0 (assume V 0) (m) EGL at Point 1, EGL1 HGL1 HV1 (m) Maximum HGL1 (m)

Min Day Avg.Day

Design Operation

Avg.Day Max Hour

Peak

0.00

0.00

0.00

0.00

0.00

0.051

0.077

0.090

0.106

0.130

103.26

103.19

103.15

103.10

103.04

103.46

103.50

103.52

103.54

103.57

0.00 103.46

0.00 103.50

0.00 103.52

0.00 103.54

0.00 103.57 103.57

Max HGL1 should equal HGL6 for aeration tank

to or after the inlet flow distribution. Good mixing should be provided to promote uniform distribution of the influent flow and RAS flow. Wastewater flow split inlet design with a relatively high headloss is often used to provide reasonably equal distribution of flow to multiple aeration tanks or to multiple inlets in each aeration tank operating in a step feed mode. Sometimes influent distribution piping which is extended to and having an inlet port at each step feed point is used. Inlet baffles. Depending on the aeration tank configuration, inlet baffles are used to dissipate the energy from the inlet velocities. Inlet baffles are designed to direct uniform distribution of MLSS along the width of the aeration tank. Aeration equipment. Diffused aeration systems are predominantly used in the municipal treatment plants. Although the air bubbles dispersed in the wastewater occupy approximately 1 percent of the volume, no allowance is made in aeration tank sizing. The volume occupied by submerged piping and diffusers is usually negligible. If spiral-flow mixing with coarse bubble diffusers is used, the width-to-depth ratios vary from 1:1 to 2.2:1. The tank depth, most commonly 4–5 m (13–16 ft), is usually determined by desired oxygen transfer efficiency of various aeration equipment. Freeboard from 0.3 to 0.6 m (1 to 2 ft) above the water surface is normally provided. If surface mechanical aerators are used, a freeboard of more than 0.6 m (2 ft) may be required depending on the power input for the aeration and mixing. Freezing during the winter due to the mist should also be considered in the design. Return activated sludge (RAS). The rate of RAS is normally 30 to 50 percent of the wastewater flow. Peak rate of RAS may go up to 100 percent of the wastewater flow for large plants and up to 150 percent of the wastewater flow for small plants. Design should provide adequate mixing, hydraulic capacity, and uniform distribution where RAS is introduced to the incoming wastewater. Effluent weir. The effluent weir provides a fixed control elevation of hydraulics in the aeration tank. Sometimes effluent ports instead of effluent weir are used to minimize headloss.




FIGURE 22.26 Schematic diagram of aeration tank system. (AT = aeration tank; PST = primary sedimentation tank).

Effluent channel. The design considerations described in the inlet channel also apply to the design of the effluent channel. Often the effluent channel from the aeration tanks is the same as the influent to the final sedimentation tanks. Hydraulic design example. The aeration tanks receive the primary sedimentation tank effluent and hydraulic loading conditions are the same as those of the primary sedimentation tanks. Design hydraulic calculations for the aeration tank system is shown in Table 22.12. The head requirements for the sample aeration tanks are in the range of 0.4–1.0 m (1.3–3.3 ft). 22.4.2.5 Granular media filter. Process criteria. Granular media filtration is usually used where the plant suspended solids effluent limit is equal to or less than 10 mg/L. It may also be applied following secondary biological treatment to remove particulate carbonaceous BOD5 and residual insolubilized phosphorus. The degree of suspended solids



22.66

Chapter Twenty-Two

removal when filtering secondary effluents without the use of chemical coagulation depends on the degree of bioflocculation achieved during secondary treatment. The presence of significant amounts of algae impedes filtration of lagoon effluents. Pretreatment with a coagulant is considered to be a good practice for such cases. There are many types of proprietary granular filters available. However, granular media filters are generally classified according to direction of flow, type, and number of media comprising the bed, the driving force, and method of flow control. Most wastewater filters are downflow units while some proprietary filters use various combinations of upflow and downflow. The driving force for filtration may be either gravity or pressure. Gravity filters are commonly used in large municipal treatment plants while pressure filters are often used in smaller plants. Gravity filters are generally sized for a filtration rate of 1.4–4 L/(m2ⴢs)/ (2–6 gal/(ft2ⴢmin) and terminal headlosses of 2.4–3.0 m (8–10 ft). Multiple units are used to allow continuous filtration during backwash or maintenance. Typical length to width ratio of gravity filters vary from 1:1 to 4:1. Key hydraulic design parameters. The key hydraulic design parameters for granular media filters include headlosses, filter operation, collection and distribution systems, and backwash requirements. Head losses. The head losses includes the losses associated with piping, valves, meters, bends, constrictions, filter media, underdrains, and collection systems. All losses vary with the square of the velocity. Clean water headloss for the filter media is influenced by media type, size, uniformity, and depth. As filtration rate increases within the terminal head loss range, less headloss capacity is available for solids storage. The head required for the filter is the sum of all headlosses including the terminal head loss of the filter media. If sufficient head is not available, pumping of filter influent is required. Filter operation. Three basic methods of filter operation are constant pressure, constant rate and variable declining rate. The constant pressure system requires a large upstream storage and is seldom used with gravity filters. The constant rate system requires a relatively costly rate control system and true constant-rate filtration is seldom used. In declining-rate filtration, the filtration rate may be kept constant using influent or effluent control weirs during the initial period of operation and, thereafter, declining rate of filtration. Generally, declining-rate filters are the best mode of gravity filter operation unless the design terminal headloss exceeds 3 m. Collection and distribution systems. (underdrain). In conventional downflow filters, the underdrain system serves to both collect the filtrate and distribute the backwash water. Traditional systems using gravel layers with perforated pipe are no longer commonly used. More popular underdrain materials include precast channels, poured-in-place concrete, or steel pipe with built-in nozzles and orifices. Porous plates made of aluminum oxide or stainless steel are also available but they are susceptible to clogging. Backwash requirements. Backwash is the cleaning of the filter by reversing the flow through the filter media at a controlled flow rate. Backwashing causes an expansion of the bed, normally no more than 10 percent of the depth, by allowing abrasive action among particles. The quantity of backwash water will generally be about 3000–4000 L/m2 (75–100 gal/ft2). Bachwashed water is collected in the wash-trough which is located about 0.9 m (3 ft) above the filter media. Biological solids in secondary effluent are strongly attached to the media and air scour before or during backwash is often required to promote successful cleaning. Air requirements for the air scour are on the order of 0.015–0.025 (m3/m2)/s [3–5 (ft3Ⲑft Ⲑ 2)/min].




Example Hydraulic Calculation of a Typical Final Aeration Tank System Initial Operation PARAMETER

1. Wastewater flow rate, Q (m3/s) (mgd) RAS flow. (% of average flow) (added downstream of aeration tank influent sluice gates) RAS flow, Qras (m3/s) 2. Aeration tanks Total of nunber of units Number of units in operation Number of units on standby Flow rate per aeration tank in operation, q (m3/s) Flow rate per aeration tank in operation including RAS flow (downstream of influent sluice gate), qras (m3/s)

Design Operation

Min Day Avg. Day Avg. Day Max Hour

Peak

1.00 23

1.60 36

2.00 46

3.20 73

3.20 73

20

50

50

100

100

0.32

0.80

1.00

2.00

2.00

3 2 1

3 2 1

3 3 0

3 3 0

3 2 1

0.50

0.80

0.67

1.07

1.60

0.66

1.20

1.00

1.73

2.60

103.57

103.57

103.57

103.57

103.57

103.67 100.67 0.66 6.00 0.15

103.67 100.67 1.20 6.00 0.23

103.67 100.67 1.00 6.00 0.20

103.67 100.67 1.73 6.00 0.29

103.67 100.67 2.60 6.00 0.38

103.82

103.90

103.87

103.96

104.05

0.03

0.04

0.03

0.05

0.07

103.85

103.94

103.91

104.01

104.12

0.66 6.0 60.0

1.20 6.0 60.0

1.00 6.0 60.0

1.73 6.0 60.0

2.60 6.0 60.0

97.87

97.87

97.87

97.87

97.87

5.95 5

6.03 5

6.00 5

6.09 5

6.18 5

Control point is located at Point 5 (aeration tank effluent weir). 3. At Point 6 Set maximum HGL6 effluent weir elevation 0.10 (m) Hydraulic Calculations Upstream of Control Point 4. Point 6 to Point 5 Headloss over sharp-crested weir Sharp-crested weir EL (m) Effluent channel bottom EL (m) Flow rate over weir, qras (m3/s) Length of weir L (m) Headloss, Hle5 (q/1.84L)(2/3) (m) HGL at Point 5, HGL5 weir EL Hle5 (m) Velocity head, HV5 V (qras/ Wp/Hle / 5)2/2g (m) EGL at Point 5, EGL5 HGL5 HV5 V (m) 5. Point 5 to Point 4 Flow rate per aeration tank in operation, qras (m3/s) Pass width, Wp (m) Tank length, Lt (m) Tank bottom elevation, ELtb Avg. day WSEL - 6 (m) Water depth in tank at design average flow, Dt (m) Number of passes per tank, Np



22.68

Chapter Twenty-Two

TABLE 22.12


Effective length of tank, L (Lt)(Np) (m) 300.0 Velocity, V V4 (m/s) 0.02 Critical depth, yc ((q2/g / / Wp2)(0.333) (m) 0.11 Friction headloss through aeration tank channel Manning’s number for concrete channel n 0.013 Hydraulic radius, R (Dt Wp)/ (2 Dt Wp) (m) 1.99 Headloss, Hlf4 f (V4 V n/R / (2/3))2 L (m) 0.0000 Fitting headloss through 90º Fitting headloss coefficient Kbend 1.0 Number of bends, Nb 8 Headloss, Hlb4 Kbend* V4 V 2/2gNb (m) 0.0001 Velocity head, Hvsd (see below at Point 3) 0.07 EGL at Point 4, EGL4 EGL5 Hlf4 f Hvsd (m) Velocity head, Hvsd (see below at Point 3) HGL at Point 4, HGL4 EGL4 HV4 V (m) 6. Point 4 to Point 3 Headloss over aeration tank influent sluice gates Sluice gate width, Ws (m) Sluice gate height (m) Flow per sluice gate, q (m3/s) Upstream head over weir, Du (select so Zsub q 0) (m) Effective sluice gate width, Ws' Ws (0.1)(2 contractions) (Du) (m) Downstream head over weir, Dd (q/1.84/Ws') (2/3) (m) Free–fall flow, Qfree 1.84 Ws'Du(3/2), (m3s) Submerged flow, Qsub Qfree (1 (Dd/ d/Du)3/2)0.385, (m3/s)

Design Operation


Peak

300.0 0.03

300.0 0.03

300.0 0.05

300.0 0.07

0.16

0.14

0.20

0.27

0.013

0.013

0.013

0.013

2.00

2.00

2.01

2.02

0.0000

0.0000

0.0000

0.0001

1.0 8

1.0 8

1.0 8

8

0.0004

0.0003

0.0009

0.0020

0.10

0.08

0.12

0.16

103.92

104.03

103.99

104.13

104.28

0.07

0.10

0.08

0.12

0.16

103.85

103.94

103.91

104.01

104.12

1.2 1.0 0.50

1.2 1.0 0.80

1.2 1.0 0.67

1.2 1.0 1.07

1.2 1.0 1.60

0.52

0.73

0.64

0.91

1.24

1.10

1.05

1.07

1.02

0.95

0.39

0.55

0.49

0.69

0.94

0.76

1.21

1.01

1.62

2.43

0.50

0.80

0.67

1.07

1.60





Design Operation


Difference, (Qsub q) (m3/s) (should be zero) 0.00 Head difference between tank and channel, Hl4 Du Dd (m) 0.13 Velocity head downstream of sluice gate, HVsd (q/Ws'/Dd / )2/2g, (m) 0.07 Velocity head upstream of sluice gate, HVsu (q/Ws'/'/Du)2/2g (m) 0.04 Top of sluice gate elevation, Els HGL4 Dd (m) 103.45 HGL upstream of sluice gate, HGLsu HGL4 Hl4 (m) 103.98 EGL upstream of sluice gate, EGLsu HGLsu HVsu (m) 104.02 Friction headloss through influent channel to tank #3 Average length of influent channel per tank, L3 31.5 Np Wp 3 tanks1/2 (m) Influent channel width, W W3 (m) 4.0 Manning’s number n for concrete channel n 0.013 Influent channel bottom elevation, Elb avg. EGLsu 3 (m) 101.1 Water depth in influent channel, h3 HGLs Elb (m) 2.87 Hydraulic radius, R (h3 w3)/ (2 h3 w3) (m) 1.18 Velocity, V3 V q/w3/h3 (m/s) 0.04 Headloss, Hlf3 f (V3 V n/R / (2/3))2 L3 (m) 0.0000 Friction headloss through influent channel to tank #2 Flow rate, q2 2 q (m3/s) 1.00 Velocity, V2 V q/w2/h2 (m/s) 0.09 Headloss, Hlf2 f (V2 V n/R / (2/3))2 L3 (m) 0.0000 Friction headloss through influent channel to tank #1 Flow rate, q1 3 q (m3/s) 1.00 Velocity, V1 q/w1/h1 (m/s) 0.09 Headloss, Hlf1 f (V1 n/R / (2/3))2 L3 (m) 0.0000

Peak

0.00

0.00

0.00

0.00

0.18

0.16

0.22

0.30

0.10

0.08

0.12

0.16

0.05

0.05

0.07

0.09

103.38

103.42

103.33

103.18

104.12

104.06

104.23

104.42

104.17

104.11

104.30

104.51

31.5

31.5

31.5

31.5

4.0 0.013

4.0 0.013

4.0 0.013

4.0 0.013

101.1

101.1

101.1

101.1

3.00

2.95

3.12

3.31

1.20 0.07

1.19 0.06

1.22 0.09

1.25 0.12

0.0000

0.0000

0.0000

0.0001

1.60 0.13

1.33 0.11

2.13 0.17

3.20 0.24

0.0001

0.0001

0.0001

0.0002

1.60 0.13

2.00 0.17

3.20 0.26

3.20 0.24

0.0001

0.0001

0.0003

0.0002



22.70

Chapter Twenty-Two

TABLE 22.12


HGL at Point 3, HGL3 HGLs Hlf3 f Hlf2 f Hlf1 f (m) 7. Point 3 to Point 2 Headloss through sluice gate Sluice gate headloss coefficient Kgate Sluice gate width, W W2 (m) Sluice gate height, Hg (m) Channel water depth, Dc (m) Gate opening depth, Hg or Dc, whichever is smaller (m) Velocity through sluice gate, V Q/W2 V5 W (m/s) Headloss, Hls2 Kgate V 2/2g (m) V5 HGL at Point 2, HGL2 HGL3 Hls2 (m)

Design Operation


Peak

103.98

104.12

104.06

104.23

104.42

1.0 1.80 1.80 2.87

1.0 1.80 1.80 3.00

1.0 1.80 1.80 2.95

1.0 1.80 1.80 3.12

1.0 1.80 1.80 3.31

1.80

1.80

1.80

1.80

1.80

0.31

0.49

0.62

0.99

0.99

0.0049

0.0124

0.0194

0.0498

0.0498

103.98

104.13

104.08

104.28

104.47

2.00 1.60 0.51 1.0

2.00 2.00 0.64 1.0

2.00 3.20 1.02 1.0

2.00 3.20 1.02 1.0

0.0132

0.0207

0.0529

0.0529

1.60 2.00 0.51 120.00 0.50 50.00

2.00 2.00 0.64 120.00 0.50 50.00

3.20 2.00 1.02 120.00 0.50 50.00

3.20 2.00 1.02 120.00 0.50 50.00

0.0001 0.0061

0.0002 0.0093

0.0004 0.0221

0.0004 0.0221

0.80 1.50 0.45 120.00 0.38 50.00

0.67 1.50 0.38 120.00 0.38 50.00

1.07 1.50 0.61 120.00 0.38 50.00

1.60 1.50 0.91 120.00 0.38 50.00

0.0001

0.0001

0.0002

0.0005

8. Point 2 to Point 1 Exit headloss from primary sed. tank effluent pipe to aeration tank influent channel Primary effluent pipe diameter, Dp (m) 2.00 All PST effluent flow, Q (m3/s) 1.00 Velocity, V1 (m/s) 0.32 Exit headloss coefficient Kexit 1.0 Exit headloss, hle1 (V12)/ 2g Kexit (m) 0.0052 Friction headloss through PST effluent pipe section 2 Flow per pipe, q (m3/s) 1.00 Pipe diameter, (Dp2) (m) 2.00 Velocity, V12 (m/s) 0.32 Hazen-Williams coefficient, Cp 120.00 Hydraulic radius, Rp2 (Dp2)/4 (m) 0.50 Length of pipe, Lp2 (m) 50.00 Slope Sp2[V12/(0.85CpRp2(0.63)](1/0.54) (%) 0.0001 Headloss, hlf2 f Lp2 Sp2 (m) 0.0026 Friction headloss through PST effluent pipe section 1 Flow per pipe, q (m3/s) 0.50 Pipe diameter, Dp1 (m) 1.50 Velocity, V11 (m/s) 0.28 Hazen-Williams coefficient, Cp 120.00 Hydraulic radius, Rp1 (Dp1)/4 (m) 0.38 Length of pipe, Lp1 (m) 50.00 Slope,Sp1[V11/(0.85CpRp1(0.63)](1/0.54) (%) 0.0001





Pipe entrance headloss Ke Headloss, hen1 Ke V112/2g (m) HGL at upstream of PST effluent pipe, HGL1 HGL2 hle1 hlf2 f hlf1 f hen1 (m) HGL7 of PST must be maximum of HGL1 (m)

Min Day Avg Day


Peak

0.50 0.0020

0.50 0.0052

0.50 0.0037

0.50 0.0094

0.50 0.0209

103.99

104.16

104.12

104.38

104.59

104.59

104.59

104.59

104.59

104.59

Hydraulic design example. A schematic diagram of a typical granular media filter system is shown in Fig. 22.27. The granular media filters receive the secondary effluent either before or after chlorination and hydraulic loading conditions are the same as those of the secondary effluent. A single granular media filter is shown for simplicity. Design hydraulic calculations for the granular media filter system is shown in Table 22.13. The head requirements for the granular media filters are in the range of 2.8–3.2 m (9.3–10.6 ft). 22.4.2.6 Mixing and contact chambers. Process criteria. Physical and chemical wastewater treatment processes involve mixing, coagulation, flocculation, and sedimentation. Chemical coagulation is often used for enhanced treatment in primary sedimentation and for tertiary treatment after secondary treatment, and before or after filtration. Advantages of coagulation include greater removal efficiencies of total suspended solids, organic materials, phosphorus, and other pollutants. Disadvantages include an increased production of chemical sludge and an increased operating cost. Chemical coagulants are mixed with wastewater during rapid mix which is the first step of the coagulation process. The coagulants destabilize the colloidal particles which allows their agglomeration. Velocity gradients (G) or a mixing intensity of 300 (mⲐ mⲐm)/s are generally sufficient for rapid mix. The rapid mix can be accomplished with mechanical mixers, in-line blenders, pumps, or air mixers. Following the rapid mixing, flocculation takes place through gentle prolonged mixing which promotes the destabilized particles to grow and agglomerate. Typical detention times for flocculation range between 20 and 30 minutes. During this period, velocity gradients of 50–80 (mⲐ mⲐm)/s should be maintained. Following flocculation, the settleable solids are settled in the following sedimentation tank. Key hydraulic design parameters. The key hydraulic design parameters for mixing and contact chambers include the inlet channel, inlet baffles, mixing equipment, and outlet channel Inlet channel. Inlet channels should be designed to maintain velocities high enough to prevent solids deposition and to promote equal distribution of flow if multiple tanks are used. Inlet baffles. Inlet baffles should be designed to dissipate the energy from the velocities and to prevent short circuiting.



22.72

Chapter Twenty-Two

Mixing equipment. A sufficient freeboard should be provided to prevent liquid spillage over the walls due to intense mixing. Provisions for easily removing the mixing equipment for repair and maintenance should be considered. Tank geometry should be configured to minimize areas with inadequate mixing. Outlet channel. Velocity in the outlet channel which leads to the sedimentation tank should be high enough to prevent solids from settling but not too high to cause breakdown of flocculated solids. 22.4.2.7 Cascade aerators Process criteria. Cascade aeration is a physical unit process typically used for effluent aeration. The system employs a series of steps or weirs over which the effluent is discharged. The system is configured to maximize turbulence in order to increase oxygen transfer. The head requirements vary depending on the initial dissolved oxygen (DO) and the desired final DO. If the necessary head is not available, effluent pumping or mechanical aeration is required. Although cascade aeration is not a new concept, its application to wastewater treatment is relatively new. Design criteria for an efficient cascade aeration system design include a fall height at each step equal to or less than 1.2 m (4 ft); a flow rate equal to or less than 235 (m3Ⲑh)/m Ⲑ [315(gal/min)/ft] of width; and a pool depth after each fall equal to or less than 0.28 m (0.9 ft). Hydraulic design example. A schematic diagram of a typical cascade aeration system is shown in Figure 22.28. Cascade aerators normally receive the secondary treatment effluent and hydraulic loading conditions are the same as those of the secondary treatment effluent. Design hydraulic calculations for the cascade aeration system is shown in Table 22.14. The head requirements for this example of the cascade aerators is 4.6 m (15.1 ft). 22.4.2.8 Effluent outfall. Process Criteria. The treatment plant accomplishes as much pollutant removal as required to produce effluent meeting the criteria established by the regulatory agencies. Ultimate disposal of wastewater effluents are by dilution in receiving waters, by discharge on land, seepage into the ground, or reclamation and reuse. Of these, disposal into the receiving waters is the most common practice. The receiving waters include rivers, lakes, estuaries, and oceans. The outfall size is determined by the velocity, headloss, structural considerations, and the economics of the situation. Velocities of 0.6–0.9 m/s (2–3 ft/s) at average flow are normally recommended in pipeline design to avoid excessive head loss. If the effluent received preliminary treatment, lower velocities can be used. However, velocities higher than 2.4–3.0 m/s (8–10 ft/s) should be avoided due to excessive headloss. Key hydraulic design parameters. The key hydraulic design parameters for effluent outfalls include available head, mixing and dispersion, submerged discharge, and diffusers. Available head. Sufficient head for gravity flow from the point of plant effluent discharge to the receiving stream is not always possible. If sufficient head is not available, effluent pumping is required to prevent flooding of the plant area. Some plants require effluent pumping during storm events or where tidal waves cause salt water intrusion. Mixing and dispersion. The outfall should be designed to operate at an adequate velocity to promote rapid dispersion and mixing of the effluent with the receiving stream. This will minimize localized deposits of settleable solids and stratification of the residual organics and nutrients in the localized area, which may cause a DO deficit and algae growth.




FIGURE 22.27 Schematic diagram of multimedia filter system. (HWL = .)

Submerged discharge. An effluent discharge pipe terminated at the bank of a stream usually leads to development of foam under low-flow conditions. The problem of foam can be overcome simply by submerging the pipe discharge below the low-water level when physical conditions in the stream allows such an arrangement. Diffusers. Certain outfalls, such as an ocean disposal, are typically accomplished by submarine outfall that consists of a long section of pipe to transport effluent and a diffuser section to dilute the effluent with the receiving stream. When the effluent water is discharged from a single- or multiport diffuser, the exit velocity will provide turbulent mixing with the surrounding water. Downloaded from Digital Engineering Library @ McGraw-Hill (www.digitalengineeringlibrary.com) Copyright © 2004 The McGraw-Hill Companies. All rights reserved. Any use is subject to the Terms of Use as given at the website.


22.74

Chapter Twenty-Two

TABLE 22.13

Example Hydraulic Calculation of a Typical Multimedia Filter System Initial Operation PARAMETER

1. Wastewater flow rate, Q (m3/s) (mgd) 2. Multimedia filters Total number of units Number of units in operation Number of units on standby Flow rate per operating multimedia filter, q (m3/s) Hydraulic Calculations at Filter Effluent 3. At Point 7 Max. HGL in filtered water storage tank, HGL7 (m) Velocity in storage tank, V V7 (m/s) Max. EGL in storage tank, EGL7 HGL7 V7 V 2/2g (m)

Min Day Avg Day


Peak

1.0 23

1.6 36

2.0 46

3.2 73

3.2 73

6 4 2

6 5 1

6 5 1

6 6 0

6 5 1

0.25

0.32

0.40

0.53

0.64

98.67 0.00

98.67 0.00

98.67 0.00

98.67 0.00

98.67 0.00

98.67

98.67

98.67

98.67

97.67

98.77 1.00 7.00 0.18 98.95

98.77 1.60 7.00 0.25 99.02

98.77 2.00 7.00 0.29 99.06

98.77 3.20 7.00 0.40 99.17

97.77 3.20 7.00 0.40 99.17

0.00

0.00

0.00

0.00

0.00

98.95

99.02

99.06

99.17

99.17

1.00 3.00 2.00 10.00 0.17

1.60 3.00 2.00 10.00 0.27

2.00 3.00 2.00 10.00 0.33

3.20 3.00 2.00 10.00 0.53

3.20 3.00 2.00 10.00 0.53

0.60 0.013

0.60 0.013

0.60 0.013

0.60 0.013

0.60 0.013

0.0001

00002

0.004

0.0009

0.0009

1.0 0.25 0.32

1.0 0.32 0.41

1.0 0.40 0.51

1.0 0.53 0.68

1.0 0.64 0.82

0.0052

0.0085

0.0132

0.0236

0.0339

98.96

99.03

99.07

99.19

99.20

4. At Point 6 Filtered water effluent channel weir Sharp-crested weir EL, Wel6 HGL7 0.1 (m) Flow rate over weir Q (m3/s) Length of weir (m) Headloss, Hlw6 (q/1.84L)(2/3) (m) HGL at Point 6, HGL6 Wel6 Hlw6 (m) Velocity in weir box, V6 (assume V 0) (m) V EGL at Point 6, EGL6 HGL6 V6 V 2/2g (m) 5. Point 6 to Point 5 Loss through effluent concrete conduit Flow rate, Q (m3/s) Width of conduit, Wc (m) Depth of conduit, Dc (m) Length of conduit, Lc (m) Velocity, Vc (m/s) Hydraulic radius, R (Wc Dc/2) /(Wc Dc) (m) Manning’s n Headloss, Hlc5 (Vc n/R / (2/3))2 Lc (m) Exit loss from pipe to concrete conduit Effluent pipe diameter, Dp (m) Pipe flow (for each filter) (m3/s) Velocity, Vp (m/s) Hle5 Vp2/2g for sharp concrete outlet (m) EGL at Point 5, EGL5 EGL6 Hlc5 Hle6 (m)





Velocity head at Point 5, HV5 V Vp2/2g (m) HGL at Point 5, HGL5 EGL5 HV5 V (m)

Design Operation


Peak

0.01

0.01

0.01

0.02

0.03

98.95

99.02

99.06

99.17

99.17

0.90

0.90

0.90

0.90

0.90

0.25

0.32

0.40

0.53

0.64

0.39 120 0.23 15.00

0.50 120 0.23 15.00

0.63 120 0.23 15.00

0.84 120 0.23 15.00

1.01 120 0.23 15.00

0.0193 0.0029

0.0305 0.0046

0.0461 0.0069

0.0785 0.0118

0.1100 0.0165

0.30 0.90

0.30 0.90

0.30 0.90

0.30 0.90

0.30 0.90

0.0024

0.0039

0.0061

0.0108

0.0155

1.20 0.32 0.0062

1.20 0.41 0.0102

1.20 0.51 0.0159

1.20 0.68 0.0283

1.20 0.82 0.0407

0.50 0.0026

0.50 0.0042

0.50 0.0066

0.50 0.0118

0.50 0.0170

98.97

99.05

99.11

99.25

99.29

0.00

0.00

0.00

0.00

0.00

98.97

99.05

99.11

99.25

99.29

2.5

2.5

2.5

2.5

2.5

101.47

101.55

101.61

101.75

101.79

0.00

0.00

0.00

0.00

0.00

6. Point 5 to Point 4 Filter effluent pipe loss Pipe diameter, Dp (m) Max. flow through filter effluent pipe q (m3/s) Velocity of flow through pipe, Vp (m/s) Hazen-Williams coefficient, Cp Hydraulic radius, Rp Dp/4 (m) Length of pipe, Lp (m) Slope, Sp[Vp/(0.85 Cp Rp0.63)](1/0.54) (%) Head loss, Hlf4 f Lp Sp (m) Headloss through butterfly valve Kvalve (fully open) Valve diameter (m) Headloss, Hval4 Kvalve (Vp2/2g) (m) Flow rate controller Venturi throat-to-inlet ratio for long tube, Krate Inlet velocity, Vi Vp (m/s) Headloss, hrate = Krate (Vi2/2g) (m) (minimum headloss when control valve is fully open) Pipe entrance loss Kent Headloss, Hlent Kent (Vp2/2g) (m) EGL at Point 4, EGL4 EGL5 Hlf4 f Hval4 Hrate Hlent (m) Velocity head, HV4 V V4 V 2/2g (assume V 0) (m) HGL at Point 4, HGL4 EGL4 HV4 V (m) 7. Point 4 to Point 3 Dirty filter head requirement, Hldf (m) (assumed) (consult with filter manufacturer) Dirty filter EGL, EGLdf HGL4 Hldf (m) Velocity head, HV3 V 0 (m) (assume V3 V 0) (m)



22.76

Chapter Twenty-Two

TABLE 22.13


Dirty filter HGL, HGLdf EGLdf HV3 V (m) 101.47 Clean filter headloss Filter bed area (m2) 160 Flow per filter, q (m3/s) 0.25 Filter rate, qfilt m3(min ⴢ m2) 0.094 Media depth, Dm (m) 1.00 Effective media size, Md (mm) 0.50 Headloss through filter, Hlf 2.32 m loss per m3(min ⴢ m2) (consultant with manufacturer) 0.2175 Entrance headloss through underdrain flume, Hlu 0.45 m m3(min ⴢ m2) 0.0422 (consult with filter manufacturer) Clean filter EGL, EGLcf EGL4 Hlf Hlu (m) 99.23 Velocity head, HV3 V 0 (assume V 0) (m) V3 Clean filter HGL, HGLcf EGLcf HV3 V (m) EGL required at Point 3, EGL3 EGLdf (m) 101.47 HGL required at Point 3, HGL3 HGLdf (m)101.47 (Head required for dirty filter controls) 8. Point 3 to Point 2 Filter inlet discharge loss Keff 1.0 Flow rate, q (m3/s) 0.25 Pipe diameter, Dp 2 (m) 0.9 Velocity, Vp2 (m/s) 0.39 Headloss, Hld2 Keff (Vp22/2g) (m) 0.0079 EGL at Point 2, EGL2 EGL3 Hld2 (m) 101.48 Velocity head, HV2 V Vp22/g / (m) 0.01 HGL at Point 2, HGL2 EGL2 HV2 V (m) 101.47 9. Point 2 to Point 1 Headloss through butterfly valve Kval (fully open) Headloss, Hlv1 Kval (Vp22/2g) Headloss through inlet pipe Length of pipe, Lp1 (m) Hazen-Williams coefficient, Cp Hydraulic radius, Rp Dp2/4 (m) Headloss, Hlf1 f (Vp2/(0.85 Cp Rp1.63)(1/0.54) Lp (m) Headloss through entrance to pipe Kent Headloss, Hlent Kent Vp2/2g (m)

Design Operation


Peak

101.55

101.61

101.75

101.79

160 0.32 0.120 1.00 0.50

160 0.40 0.150 1.00 0.50

160 0.53 0.200 1.00 0.50

160 0.64 0.240 1.00 0.50

0.2784

0.3480

0.4640

0.5568

0.0540

0.0675

0.0900

0.1080

99.38

99.52

99.81

99.95

101.55 101.55

101.61 101.61

101.75 101.75

101.79 101.79

1.0 0.32 0.9 0.50 0.0129 101.56 0.01 101.55

1.0 0.40 0.9 0.63 0.0202 101.63 0.02 101.61

1.0 0.53 0.9 0.84 0.0359 101.79 0.04 101.75

1.0 0.64 0.9 1.01 0.0517 101.084 0.05 101.79

0.3 0.0024

0.3 0.0039

0.3 0.0061

0.3 0.0108

0.3 1.0155

20.0 120 0.23

20.0 120 0.23

20.0 120 0.23

20.0 120 0.23

20.0 120 0.23

0.0039

0.0061

0.0092

0.0157

0.0220

0.50 0.0039

0.50 0.0065

0.50 0.0101

0.50 0.0179

0.50 0.0258





EGL at Point 1, EGL1 EGL2 Hlv1 Hlf Hlent (m) Velocity head, HV1 0 (assume V1 0) (m) HGL at Point 1, HGL1 EGL1 HV1 (m) Minimum required control HGL at Point 1 (m) (Max. HGL1 must equal HGL7 of final sedimentation tank)

Design Operation


Peak

101.49

101.58

101.65

101.83

101.90

0.00

0.00

0.00

0.00

0.00

101.49

101.58

101.65

101.83

101.90

101.90

101.90

101.90

101.90

101.90

22.4.2.9 Slurry and chemical pumping. Sludge solids. Typical needs for sludge pumping involve transporting sludge from primary and secondary clarifiers to and between thickening, conditioning, digestion or dewatering facilities, and from biological processes for recycle or further treatment. Several different types of sludge pumps are used since various types of sludge require a wide range of service conditions. The flow characteristics (rheology) of wastewater sludges vary widely from process to process and from plant to plant. Because rheological properties directly influence pipeline friction losses of pumped sludges, head loss characteristics of wastewater sludges also vary extensively. Minimizing pumping distance and applying a conservative multiplier to headlosses calculated for equivalent flows of water is the traditional approach to the design of sludge pumping and piping systems. However, this approach is often inadequate. As a result of past research of non Newtonian fluid characteristics of sludges, sludge pumping system design data based on specific measured rheological characteris-

FIGURE 22.28 Schematic diagram of cascade aeration system.



22.78

Chapter Twenty-Two

tics of sludge and the characteristics on piping systems are now available. These data are presented in Section 22.5. Scum. Scum is collected from the surface of primary sedimentation tank or secondary sedimentation tank. Scum from the secondary treatment is more dilute and is usually returned to the head of the treatment plant or thickened prior to combining the thickened scum with that from primary treatment. The scum is collected to a scum wet well and pumped to another location for processing. Progressive cavity pumps, pneumatic ejectors, and recessed impeller centrifugal pumps are used to pump scum. Key design elements for the scum collection and handling system include sloping the bottom of the scum

TABLE 22.14

Example Hydraulic Calculation of a Typical Cascade Aeration System Initial Operation PARAMETER

1. Wastewater flow rate, Q (m3/s) (mgd) 2. Cascade aerator Total number of units Flow rate through aerator, Q (m3/s) Optimal flow rate per m width over step, q (m3/s) DO concentration of postaeration influent, CO (mg/L) Desired DO concentration of postaeration effluent, Cu (mg/L) Calculation of aerator dimensions with predetermined weir length 3. Weir length, W (m) Flow over weir, q Q/W, W (m3/s) Critical depth at upstream step edge, hc (q2/g / )1/3 (m) Optimal fall height of nappe, h, Length of downstream bubble cushion, Lo 0.0629(h0.134)(q0.666) (m) Length of downstream receiving channel, L 0.8Lo (m) Optimal tailwater depth, H' 0.236h (m) for h 1.2 m Deficit ratio log at 20º C, In(r20) 5.39(h1.31)(q0.363)(H H0.31) Deficit ratio, r20

Design Operation


Peak

1.0 23

1.6 36

2.0 46

3.2 73

3.2 73

1 1.00

1 1.60

1 2.00

1 3.20

1 3.20

0.0653

0.0653

0.0653

0.0653

0.0653

0.00 5.00

0.00 5.00

0.00 5.00

0.00 5.00

0.00 5.00

5.0

5.0

5.0

5.0

5.0

0.20

0.32

0.40

0.64

0.64

0.160 1.2

1.219 1.2

0.254 1.2

0.347 1.2

0.347 1.2

5.16

7.05

8.18

11.19

11.19

4.12

5.64

6.54

8.95

8.95

0.28

0.28

0.28

0.28

0.28

0.42 1.53

0.36 1.43

0.33 1.39

0.28 1.32

0.28 1.32

Calculate concentration of dissolved oxygen downstream of step. If concentration is less than desired downstream concentration, add another step and again calculate DO downstream concentration. Continue adding steps until the desired DO concentration is achieved.





Design Operation


Peak

Select cascade aerator dimensions corresponding to those calculated for average flow. 4. Calculation of number of steps to obtain desired DO Desired DO concentration at average flow, Cu (mg/L) Step 1 effluent DO, C1 9.07 (1 (1/r20)) CO/r20) (mg/L) Step 2 effluent DO, C2 9.07 (1 (1/r20)) C1/r20) (mg/L) Step 3 effluent DO, C3 9.07 (1 (1/r20)) C2/r20) (mg/L)

5.00 3.14

2.73

2.55

2.21

2.21

4.81

4.51

4.39

4.14

4.14

6.01

5.80

5.70

5.52

5.52

1.00 1.20

1.00 1.20

1.00 1.20

1.00 1.20

1.00 1.20

HGL at Point 1, HGL1 (m)

97.53

97.53

97.53

97.53

97.53

HGL at Point 2, HGL2 HGL1 h (m) HGL at Point 3, HGL3 HGL2 h (m) HGL at Point 4, HGL4 HGL3 h (m)

96.33 95.13 93.93

96.33 95.13 93.93

96.33 95.13 93.93

96.33 95.13 93.93

96.33 95.13 93.93

In this example, the desired downstream DO concentration for average flow is achieved after three steps. 5. Calculation of HGL at each step Head loss from filtered water storage tank to point 1 (m) Cascade fall height, h (m)

tank, use of smooth pipe such as glass-lined pipe, providing flushing connections, pigging stations and cleanouts. Grit slurry. Removal and conveyance of grit from the grit chamber can be accomplished with varying degrees of success by a number of different methods, including inclined screw or tubular conveyers, chain and bucket elevators, clamshell buckets, and pumping. Of these methods, pumping of grit from hoppers in the form of slurry offers distinct advantages over other methods but also has some disadvantages. The advantages include small space requirement and flexibility of service by any grit pump from any grit tank to any grit handling system with simple valve operation. A disadvantage is frequent maintenance required for piping and valves due to the abrasive grit. Considerations to be given in piping design include minimization of bends, providing cleanouts at critical bends, providing redundant piping at the location of likely clogging, and maintaining a velocity of 1–2 m/s (3–6 ft/s). Vortex or recessed impeller pumps and air lift pumps normally handle grit slurries. Frequent pumping and applying waterjets or compressed air to loosen the compacted grit in the hopper prior to pumping is a good practice for grit pumping. Chemical solutions. Chemicals used in municipal treatment plants are received in either liquid or solid form. The chemicals in solid form generally are converted to soluDownloaded from Digital Engineering Library @ McGraw-Hill (www.digitalengineeringlibrary.com) Copyright © 2004 The McGraw-Hill Companies. All rights reserved. Any use is subject to the Terms of Use as given at the website.


22.80

Chapter Twenty-Two

tion or slurry prior to feeding although dry feeding is also practiced. Design of solution feed systems mainly depends on liquid volume and viscosity. Liquid feed units include piston, positive-displacement, and diaphragm pumps, as well as liquid gravity feeders. The unit best suitable for a particular application depends on the required head, chemical corrosiveness, application rate, other liquid properties, and the type of control.

22.5 NON-NEWTONIAN FLOW CONSIDERATIONS This section addresses pipe transport of mixtures of solids in a liquid media. This is relevant to us for the analysis of wastewater sludge transport. When a fluid motion begins within a pipe, the velocities of flow at all points along the cross section of the pipe are equal. Over time, velocity gradients are established, beginning at the wall of the pipe due to the resistance forces developed at the fluid-solid interface. Eventually the velocity gradients extend throughout the cross section of the flow. The velocity gradients result from the relative movement between fluid layers and the resultant shear. Fluids resist shear and, therefore, shear stresses are caused within a fluid in motion in a pipe. For water and other newtonian fluids, the shear stress is directly proportional to the velocity gradient. Many suspensions behave in non-newtonian fashion, as plastic fluids. In thin suspensions, the suspended particles are not in contact and the suspension will exhibit the newtonian properties of water. When the concentration becomes sufficiently great to force the particles into contact with each other, a measurable stress is needed to produce motion. Experiments by Bingham (1922) and Babbitt and Caldwell (1939) demonstrated that sewage sludges exhibit both types of flow characteristics depending on the type of solids and the moisture content. At low solid concentrations, the solid particles are generally not in contact with one another. In this case the presence of the solids has negligible impact on the density and the viscosity of the liquid. As the solids concentration increases, the suspended particles come into contact with each other and the resultant shearing stress must be overcome before any movement can start. Under such conditions, the flow assumes plastic characteristics and the headloss varies almost directly with the reduction of moisture M. The headlosses associated with the two types of flow are different. The dividing point between these two is called the limiting moisture content ML, which is defined as the moisture content in percent where a measurable yield stress, Sy, first occurs. As described by Chou (1958), below ML, the flow is plastic, and, above it, the flow is in suspension only. Furthermore, it is generally recognized that in sludge flow, as in other fluid flow, there is a critical velocity and, consequently, the Reynolds number, which divides the flow into laminar and turbulent stages. With flow in suspension there is no yield stress value and the Reynolds number takes the form of ρV VD Re (22.2) µ where Re Reynolds number ρ specific weight, V velocity, D pipe diameter, µ coefficient of viscosity similar to that for water. In plastic flow the apparent viscosity decreases with the increase in velocity, as discussed by Hatfield (1938) and, in a given range, it may be expressed as 16Sy D µ η 3V

(22.3)




where Sy yield stress η coefficient of rigidity and the corresponding Reynolds number becomes ρV VD 3ρDv2 R 3ηv 16SyD µ

(22.4)

Using the Babbitt and Caldwell (1939) recommendation, and taking 2000 as the lower and 3000 as the upper limits of R, the critical velocities are: For flow in suspension, where M ML (flow in suspension); 2000µ VLC ρD

(22.5)

3000µ VUC ρD

(22.6)

For plastic flow, where M ML (plastic flow): 1000η 103 兹兹9 苶苶4苶 η2苶苶苶 D2苶S苶 yρ VLC ρD

(22.7)

η2苶苶苶 D2苶S苶 1500η 127 兹兹1苶苶4苶0苶 yρ VUC ρD

(22.8)

where VLC lower critical velocity VUC upper critical velocity The yield stress value Sy, the coefficient of rigidity η, and the specific weight are the basic variables required in computing critical velocities and headlosses. These properties vary from sludge to sludge depending on characteristics such as moisture M, nature of the suspended particles, temperature, and extent of turbulence. These factors also influence each other making it difficult to develop a useful equation for engineering practice. To resolve this issue, Chou presented an approach using moisture content M as the principal index of sludge, while placing all other parameters into the general term “origin or kind of sludge,” such as “”primary,” “digested,” and “digested from Imhoff tank,” and so on. The following development was presented by Chou (1958). The graphical values were taken from Babbitt and Caldwell (1938) and Keefer (1940). G related to M. Specific gravity, G, is primarily used in computing specific weight as in ρ 62.4 G. Specific gravity for activated sludge was shown to be, G 1.007, at a typical moisture content of about 98 to 99 percent. Primary sludges are more variable, but the curve in Fig. 22.29 indicates a reasonable mean. In Figure 22.29, digested sludges have a cluster of points near G 1.025, but the curve shows the general tendency. The three points from Imhoff tanks are on a smooth curve. Sy related to M. Yield stress, Sy, has an important role to play in calculating headloss and critical velocities. In Fig. 22.30 the two curves marked with “Imhoff Tank” and “Good Digestion” were considered to be representative of true conditions. The “Primary” values based on two points are clearly an approximation. The rest of the points varied consider-



22.82

Chapter Twenty-Two

ably and were designated as the curve of “Poor Digestion” in an attempt to represent the upper limit of the range. The limiting moisture ML can be determined from Fig. 22.30 as the point where Sy 0 cuts the curve. η related to M. The experimental determination of the coefficient of rigidity indicated its variation with moisture M to be less pronounced than Sy. Accordingly, the plots are more scattered. The two lines (shown in Fig. 22.31) of “High” and “Mean” are suggested for design purposes. Case 1—Suspension/Laminar Stage. For flow in suspension, the solid particles are free to move past one another and there is consequently no yield value to overcome. Reduction of moisture content only slightly increases the specific weight ρ (ρ 62.4 G) and the viscosity µ. Both remain close to the values for water. The yield stress, Sy, is zero for flow in suspension. The equation for headloss for laminar stage flow in suspension becomes ηV H 2 L 62.4G D

(22.9)

where G specific gravity in which both G and η for the corresponding M can be determined from the Figs. 22.29 and 22.31. Case 2—Suspension/Turbulent Stage. Streck (1950) and Winkel (1943) reported the headloss of turbulent flow in suspension may be computed as follows: HS G2HW

(22.10)

where HS the headloss of flow in suspension with moisture M HW the corresponding headloss of pure water G the specific gravity of the suspension (from Fig. 22.29) The headloss of flow in suspension for both laminar and turbulent conditions is not significantly greater than the corresponding headloss for water. Case 3—Plastic Flow/Laminar Stage. Plastic flow in the laminar stage is the most common case in sludge flow. As discussed above, the headloss is partly due to yield value and partly due to coefficient of rigidity, both of which are affected by the moisture M. Babbitt and Caldwell (1939) reported headloss for this case as follows: ηV H 16Sy 2 L 3ρD ρD

(22.11)

in which the values of ρ, Sy, and η may be determined from Figs. 22.29, 22.30 and 22.31, respectively. For any moisture below the limiting value, plastic flow conditions mean Sy

0 and a headloss occurs due to yield value, Sy, alone. As motion begins, headloss increases with the first power of velocity in the laminar stage. Hence, as soon as the applied head is greater than Sy, relatively little additional head is required to accelerate the flow to critical velocity. Therefore, it may be concluded that the most economical velocity of sludge flow is the critical velocity, above which the headloss increases rapidly with the velocity. Case 4—Plastic Flow/Turbulent Stage. Published data for turbulent plastic flow headloss are variable and inconsistent. Due to variation of sludge characteristics, the velocities, the results are extremely unpredictable.




For fully turbulent flow, it seems reasonable that the headloss results primarily from kinetics and is proportional to v2/2g and the specific weight ρ and, therefore, will differ from that of water only slightly by the effect of ρ. This ideal condition of full turbulence rarely occurs for plastic flows. As the moisture drops below ML, the critical velocities increase and the thickness of the boundary layers is increased in proportion to moisture reduction. The velocity distribution in a cross section and the impacts of the boundary layers are not the same as the regular patterns of homogeneous liquids. Due to the complicated and variable phenomena occurring during turbulent plastic flow, it is difficult, if not impossible, to accurately anticipate headloss for flow in this condition. Designing for this

SPECIFIC GRAVITY G

FIGURE 22.29 Specific gravity G of sludge (From Chou, 1958)



22.84

Chapter Twenty-Two

condition is uncertain and not recommended. However, some experimental data are available for guidance when turbulent plastic flow is unavoidable. Brisbin (1957) compiled headloss data for raw, thickened sludge. Thus, from such complicated phenomena, uniform results can hardly be expected. The corresponding C in the Hazen-Williams formula V 1.318Cr0.63 s0.54

(22.12)

where r hydraulic radius and s H/ H/L hydraulic slope was computed from the observed headlosses. These C ' values are tabulated in Table 22.15 along with the ratio to water headloss.

Yield Stress, Sy FIGURE 22.30 Yield value of Sy of sewage sludges (From Chou, 1958)




22.5.1 Headloss Computation With the source and M of the sludge known or assumed, the first step is to determine if the flow is a suspension or plastic. Empirically this can be done by the curves in Fig. 22.30. Values for G, Sy and η are then chosen from curves in Figs. 22.29, 22.30, and 22.31. Example. Given primary sludge, M 95. The flow is plastic since M ML (M ML 99.8 percent at point in Fig. 22.30 where Sy 0). From Figs. 22.29, 22.30 and 22.31, G 1.022, 1.022 62.4 63.77 lb/ft2 Sy 0.065 lb/ft/s η 0.0127 (lbⴢft)/s Critical velocities (22.13)

PERCENTAGE OF MOISTURE BY WEIGHT

103兹0 兹苶苶 .0苶1苶5苶1苶3苶苶4苶 .1苶4苶5苶 D2苶 VLC 12.7 63.77D

FIGURE 22.31 Coefficient of rigidity n of sludge (From Chou, 1958)



22.86

Chapter Twenty-Two

127兹0 兹苶苶 .0苶2苶2苶6苶苶4苶 .1苶4苶5苶 D2苶 VUC 19.1 63.77D

(22.14)

The values are given in Table 22.16. Laminar stage H 0.00555 V 0.000204 2 L D D

(22.15)

The values are tabulated against the pipe diameter D for a range of laminar flow velocities in Table 22.17. Turbulent stage: Assume C 100 for M 100, and from a plot of Table 1 C' values, the corresponding C' 54.7 for M 95. V 72.09r 0.63s0.54

(22.16)

V1.85 V1.85 s H L 72.091.85r1.165 Constant The headlosses are computed in Table 22.18. It is useful to plot results as shown in Figs. 22.32 and 22.33 with critical velocities indicated. For laminar flow, values are taken from the left of VLC, and for turbulent flow, they are taken from right of VLC. It is also useful to tabulate results as shown in Table 22.19, including the minimum headloss to account for Sy as well as the operating headloss. Head losscomputations for solids bearing flows are not an exact science. Where the physical properties of the sludge cannot be measured, use of the data reproduced here in Figs. 22.29, through 22.31 and the methodology developed by Chou et al. (1958) and summarized here should provide reasonable results.

TABLE 22.15

C’ Values for Raw, Thickened Sludge

M

C’

Moisture Content (%)

Percentage of C at M 100%

Ratio to Water Headloss  100 1.85   C’ 

100

100

100

98

80.5

1.49

97

—

—

96

62.8

2.37

95

—

—

94

50.5

3.54

91.5

37.6

6.11

90

33.6

7.54




FIGURE 22.32 Results of Headloss computation examples–laminar flow



22.88

Chapter Twenty-Two

FIGURE 22.33 Head loss for turbulent flow (m 95%)




Example of Critical Velocities

D

8 in

10 in

14 in

20 in

VLC (ft/s) VUC (ft/s)

3.58

3.55

3.45

3.40

4.52

4.42

4.31

4.23

10 in

14 in

20 in

0.000293v

0.000149v

0.000073v

0.00666

0.00476

0.00333

0.00754

0.00521

0.00355

0.00770

0.00527

0.00358

TABLE 22.17

Example Hydraulic Slope for Laminar Stage

D

8 in

H  ft   0.000458v L  ft  H V 0, 0.00833 L V 3, H 0.00970 L VLC , H 0.00997 L Varies (see Table 22.16)

TABLE 22.18

Example Hydraulic Slope for Turbulent Stage D:8 in

D:10 in

D:14 in

D:20 in

V

V1.85 338.9

V1.85 439.6

V1.85 650.5

V1.85 986.5

Varies Table 22.16 19.64 27.51 36.60 46.85 58.25 70.80

0.0481 0.0580 0.0813 0.108 0.138 0.172 0.209

0.0356 0.0447 0.0625 0.0832 0.106 0.133 0.161

0.0229 0.0302 0.0423 0.0562 0.072 0.0896 0.109

0.0146 0.0199 0.0279 0.0371 0.0474 0.0591 0.0718

gal/m

ft3/s

ft/s

2000 2000 4000 4000 600 500 1600 3300

4.46 4046 8.91 8.91 1.34 1.11 3.57 7.35

8.20 12.80 8.36 4.08 2.44 3.21 3.34 3.37

1.85

V fps VUC 5 6 7 8 9 10

TABLE 22.19 Summary of Results Pipes L ---------- D 16 ft ------- 0 in 5 ft ------- 8 in 11 ft ------- 14 in 20 ft ------- 20 in 20 ft ------- 10 in 50 ft ------- 8 in 30 ft ------- 14 in 40 ft ------- 20 in

Q

Headloss, feet Minimum(1) Operating 0.11 0.04 0.05 0.08 0.13 0.42 0.14 0.13

1.78 1.65 0.86 0.28 0.15 0.49 0.16 0.14

Minimum is the headloss required to overcome Sy and initiate flow.



22.90

Chapter Twenty-Two

REFERENCES American Society of Civil Engineers, and Water Environment Federation, Gravity Sanitary Sewer Design and Construction, American Society Civil Engineers Manuals and Reports on Engineering Practice No. 60 and Water Environment Federation Manual of Practice No. FD-5, 1982. Babbitt, H. E., and Caldwell, David H., Laminar Flow of Sludges in Pipes with Special Reference to Sewage Sludge, University of Illinois, Bulletin 319, 1939. Bingham, E. C., Fluidity and Plasticity, McGraw-Hill, New York, 1922. Brisbin, S. G., “Flow of Concentrated Raw Sewage Sludges in Pipes,” Proceedings Paper 1274, American Society Civil Engineers 1957. Bulletin No. 2552 University of Wisconsin. Bureau of Reclamation, Design Standards No.3, Water Conveyance Systems, Chapter 11 General Hydraulic Considerations (Draft), (7-2071) (6-84), Sept. 30, 1992. Camp, T. R., and Graber, S. D., Dispersion Conduits, Journal of the Sanitary Engineering Division, American Society of Civil Engineer, 94(SA1), February 1968. Chao, J.–L., and Trussell, R. R., “Hydraulic Design of Flow in Distribution Channels,” Journal of Environmental Engineering Division, ASCE, 6(EE2), April 1980. Chou, T.–L., “Resistance of Sewage Sludge to Flow in Pipes,” Journal of Sanitary Engineering Div., American Society of Civil Engineer, Paper 1780, September 1958. Committee on Pipeline Planning, Pipeline Division, Pipeline Design for Water and Wastewater, American Society of Civil Engineers, New York, 1975. Crane Co., “Flow of Fluids Through Valves, Fittings, and Pipe”, Technical Paper No. 410-C, 23rd ed., Banford, Ontario, 1987. Daugherty, R. L., and J. B. Franzini, Fluid Mechanics with Engineering Applications, 7th ed., McGraw-Hill, New York, 1977. Hatfield, W. D., “Viscosity or Psendo-Plastic Properties of Sewage Sludges,” Sewage Works Journal, 10, 1938. Ito, H., and Imani, K., “Energy Losses at 90o Pipe Junctions.” Journal of the Hydraulics Division, American Society of Civil Engineer, HY9, 1973. Keefer, C. E., Sewage Treatment Works, McGraw-Hill, New York, 1940. Sanks, R. L., Pumping Station Design, Butterworths, Stoneham, MA, 1989. Shaw, G. V., and A. W. Loomis, eds., Cameron Hydraulic Data, Ingersoll-Rand Co., Cameron Pump Division, 14th Ed., 1970. Simon, A. L., Hydraulics, 3rd ed., John Wiley & Sons, New York, 1986. Streck, O., Grund und Wasserbrau in Praktischen Biespielen, Springer-Verlag, Berlin, 1950. Ten-State Standards, Recommended Standards for Sewage Works, Great Lakes–Upper Mississippi Board of Sanitary Engineers, Health Education Service, Inc., Albany, NY, 1978. Walski, T. M., Analysis of Water Distribution Systems, Krieger, Malabar, FL, 1992. Williamson, J. V., and Rhone, T. J., ßDividing Flow in Branches and Wyes,” Journal of the Hydraulics Division, American Society of Civil Engineer, No. HY5, 1973. Winkel, R., Angwandte Hydromechanik im Wasserbau, Ernst & Sohn, Berlin, 1943. Yao, K. M., Hydraulic Control for Flow Distribution, Journal of the Sanitary Engineering Division, American Society of Civil Engineer, 98 (SA2), April 1972




APPENDIX

WATER AND WASTEWATER TREATMENT PLANT HYDRAULICS ENGLISH UNITS EXAMPLES

TABLE 22.5

Hydraulic Calculations of a Typical Coagulation Process Initial Operation Parameter

1. Plant Flow (mgd) (ft3/s)

Design Operation

Min Day

Avg Day

Avg Day Max Hour

50 77.36

70 108.31

75 116.04

100 154.72

360.1

360.01

360.02

360.04

50.77 10.26 172.72 0.29 4.62

71.08 10.27 172.89 0.41 4.63

76.15 10.27 172.94 0.44 4.63

101.54 10.29 173.25 0.59 4.63

0.00 360.01

0.00 360.02

0.00 360.02

0.00 360.04

24.18 10.26 172.72 0.14 4.62

33.85 10.27 172.89 0.20 4.63

36.26 10.27 172.94 0.21 4.63

48.35 10.29 173.25 0.28 4.63

0.00 360.01

0.00 360.02

0.00 360.02

0.00 360.04

9.67 10.26 172.72 0.06 4.62

13.54 10.27 172.89 0.08 4.63

14.51 10.27 172.94 0.08 4.63

19.34 10.29 173.25 0.011 4.63

Note: For Points 1 through 8, see Fig. 22.12. 2. WSEL at Point 1 (calculation done in Table 22.6) (ft) 3. Point 1 to Point 2 Average Flow 21Q/32 (ft3/s) Flow depth WSEL @ 1 invert (349 ft 9 in) (ft) Flow area 16ft 10in width depth (ft2) Velocity flow/area (ft/s) R = A/P / (P w + 2d) (ft) 2 Condiut loss [(V n )/(1.486 R2/3)] L (ft) where n 0.014 and L 95 ft WSEL at Point 2 (ft) 4. Point 2 to Point 3 Average flow 5Q/16 (ft3/s) Flow depth = WSEL @ 2 – invert (349 ft 9 in) (ft) Flow area 16 ft – 10 in width depth (ft2) Velocity flow/area (ft/s) R = A/P / (P w 2d) (ft) 2 Condiut loss [(V n)/(1.486b R2/3) ] L (ft) x L where n 0.014 and L 48 ft WSEL at Point 3 (ft) 5. Point 3 to Point 4 Average flow Q/8 (ft3/s) Flow depth WSEL @ 3 – invert (349 ft – 9 in) (ft) Flow area 16 – ft 10 in width depth (ft2) Velocity flow/area (ft/s) R= A/P / (P w 2d) (ft)



22.92

Chapter Twenty-Two

TABLE 22.5


Condiut Loss [(V n)/(1.486 R2/3)]2 L (ft) where n 0.014 and L 72 ft WSEL at Point 4 (ft) 6. Point 4 to Point 5 Flow Q/32 (ft3/s) Port area 1 ft deep 2.5 ft wide (ft2) Velocity flow/area (ft/s) Submerged entrance loss 0.8 V 2/2g (ft) WSEL at Point 5 (in sedimentation tank) (ft) 7. Point 5 to Point 6 Width of sedimentation basin (W) W (ft) Flow (Q/4) (ft3/s) Invert elevation of sedimentation baffles (ft) Fow depth (H) H (WSEL at Point 5 – baffle invert) (ft) Area downstreams of baffle (W H H) (ft2) Horizontal openings in baffles, 1 in wide, every 9 inches Area of openings, A W H H/ 9 (ft2) Velocity of downstream baffle (V downstream) (Q/A) (ft/s) Velocity of 1 in opening section (V1) (Q/A / ) (ft/s) Local losses Sudden expansion (1.0 V downstream2/2g) sudden contraction (0.36 V12/2g) (ft) WSEL at Point 6 (Upstream of sedimentation baffles) (ft)

Design Operation

Min Day

Avg Day

Avg Day Max Hour

0.00 360.01

0.00 360.02

0.00 360.02

0.00 360.04

2.42 2.50 0.97 0.01 360.02

3.38 2.50 1.35 0.02 360.04

3.63 2.50 1.45 0.03 360.05

4.84 2.50 1.93 0.05 360.09

76.00 19.34 347.67 12.35 938.77

76.00 27.08 347.67 12.37 940.39

76.00 29.01 347.67 12.38 940.88

76.00 38.68 347.67 12.42 943.83

104.31

104.49

104.54

104.87

0.02

0.03

0.03

0.04

0.19

0.26

0.28

0.37

0.00

0.00

0.00

0.00

360.02

360.04

360.05

360.09

8. Point 6 to Point 7 Loss per stage (provided by flocculator manufacturer) (ft) Total loss (three stages) (ft) WSEL at Point 7 (ft)

0.04 0.13 360.15

0.04 0.13 360.17

0.10 0.29 360.34

0.17 0.50 360.59

9. Point 7 to Point 8 Flow Q/24 (ft3/s) Port area 1 in deep 1 ft – 6 in wide (ft2) Velocity flow/area (ft/s) Entrance loss 1.25 V 2/2g (ft) WSEL at Point 8 (inlet port) (ft)

3.22 1.50 2.15 0.09 360.24

4.51 1.50 3.01 0.18 360.35

4.84 1.50 3.22 0.20 360.54

6.45 1.50 4.30 0.36 360.95

3.22 2.24 6.73 0.48 0.90

4.51 2.35 7.05 0.64 0.92

4.84 2.54 7.63 0.63 0.94

6.45 2.95 8.84 0.73 0.99

Note: For Points 8 through 14, see Fig. 22.13 10. Point 8 to Point 9 Average Flow Q/24 (ft3/s) Flow depth WSEL @ 8 - invert (358 ft) (ft) Flow area 3 ft width depth (ft2) Velocity flow/area (ft/s) R = A/P / (P w 2d) (ft)





Min Day

Avg Day


2

Condiut loss [(V n)/(1.486 R 2/3)] L where n = 0.014 and L 12 ft – 8in (ft) WSEL at Point 9 (ft)

0.00

0.00

0.00

0.00

360.24

360.35

360.54

360.95

11. Point 9 to Point 10 Average flow Q/12 (ft3s) Flow depth WSEL @ 9 – invert (358 ft) (ft) Flow area 3 ft width depth (ft2) Velocity flow/area (ft/s) R = A/P / (P w 2d) (ft) Condiut loss [(V n)/(1.486 R 2/3)]2 L (ft) where n 0.014 and L 12 ft – 8 in (ft) WSEL at Point 10 (ft)

6.45 2.24 6.73 0.96 0.90

9.03 2.35 7.05 1.28 0.92

9.67 2.54 7.63 1.27 0.94

12.89 2.95 8.85 1.46 0.99

0.00 360.24

0.00 360.35

0.00 360.54

0.00 360.95

12. Point 10 to Point 11 Flow Q/8 (ft3/s) Flow depth WSEL @ 10 – invert (358 ft) (ft) Flow area 3 ft width depth (ft2) Velocity flow/area (ft/s) Loss at two 45o bends 2 0.2 V 2/2g (ft) WSEL at Point 11 (ft)

9.67 2.24 6.73 1.44 0.01 360.26

13.54 2.35 7.06 1.92 0.02 360.38

14.51 2.54 7.63 1.90 0.02 360.57

19.34 2.95 8.85 2.18 0.03 360.98

13. Point 11 to Point 12 Flow Q/4 (ft3/s) Flow depth WSEL @ 11 – invert (358 ft) (ft) Flow area 5 ft width depth (ft2) Velocity flow/area (ft/s) Loss at two 45o bends 2 0.2 V2/2 V g (ft) R A/P / (P w 2d) (ft) Condiut Loss [(V n)/(1.486 R 2/3)]2 L (ft) where n 0.014 and L 32 ft WSEL at Point 12 (ft)

19.34 2.26 11.28 1.71 0.02 1.19

27.08 2.38 11.88 2.28 0.03 1.22

29.01 2.57 12.83 2.26 0.03 1.27

38.68 2.98 14.90 2.60 0.04 1.36

0.01 360.28

0.01 360.42

0.01 360.61

0.01 361.04

14. Point 12 to Point 13 Flow Q/4 (ft3/s) Flow depth WSEL @ 12 – invert (358 ft) (ft) Inlet area 5 ft width depth (ft2) Velocity flow/area (ft/s) Inlet loss 1 V 2/2g (ft) WSEL at Point 13 (Mixing chamber No. 2 outlet) (ft)

19.34 2.28 11.41 1.70 0.04 360.33

27.08 2.42 12.09 2.24 0.08 360.50

29.01 2.61 13.05 2.22 0.08 360.69

38.68 3.04 15.18 2.55 0.10 361.14

19.34 36.00 0.54

27.08 36.00 0.75

29.01 36.00 0.81

38.68 36.00 1.07

0.01

0.02

0.03

0.05

360.34

360.52

360.71

361.19

15. Point 13 to Point 14 Note: Mixers provide negligible head loss Flow Q/4 (ft3/s) Chamber area 6 ft 6 ft (ft2) Velocity flow/area (ft/s) Losses Mixer (1 V 2/2g) Sharp bend (1.8 V 2/2g) (ft) WSEL at Point 14 (Mixing Chamber No. 2 inlet) (ft)



22.94

Chapter Twenty-Two

TABLE 22.5


Design Operation

Min Day

Avg Day

Avg Day Max Hour

38.68 30.00 1.29 1.30

54.15 30.00 1.81 1.30

58.02 30.00 1.93 1.30

77.36 30.00 2.58 1.30

0.02

0.03

0.03

0.06

0.02 360.37

0.03 360.58

0.04 360.79

0.07 361.31

360.31

360.51

360.70

361.17

360.34

360.54

360.74

361.24

77.36 30.00 2.58 1.30

108.31 30.00 3.61 1.30

116.04 30.00 3.87 1.30

152.72 30.00 5.16 1.30

0.15 360.49

0.28 360.82

0.31 361.06

0.53 361.77

77.36 30.00 30.25 30.13 2.57 1.30 1.38 1.34

108.31 30.00 30.25 30.13 3.60 1.30 1.38 1.34

116.04 30.00 30.25 30.13 3.85 1.30 1.38 1.34

154.72 30.00 30.25 30.13 5.14 1.30 1.38 1.34

0.01 360.50

0.02 360.84

0.02 361.08

0.04 361.81

Note: For Points 14 through 21, see Fig. 22.14 16. Point 14 to Point 15 Flow Q/4 (ft3/s) Condiut area 7.5 ft wide 4 ft deep (ft3) Velocity flow/area (ft/s) R A/P / (P 2w 2d) (ft) Condiut losses L [V/(1.318 V C R0.63)]1/0.54 (ft) where L 155 ft and Hazen-Williams C 120 Local losses Flow split (0.6 V 2/2g) contraction (0.07 V 2/2g) 0.67 V 2/2g (ft) WSEL at Point 15 (at Mixing Chamber No. 1) (ft) 17. The above calculations (for Points 1 through 15) have been routed through Tank No. 4 When the flow isrouted through Tank No. 1, the WSEL (ft) is: In reality, the headloss through each basin is equal. The flow through the basin naturally adjusts to equalize headlosses, that is flow through Tank No. 1 is greater than Q/4 and flow through Tank No. 4 is less than Q/4. The actual headloss through each basin is the average of Tank #’s 1 and 4 and the WSEL (ft) at Point 15 is: 18. Point 15 to Point 16 Flow Q (ft3/s) Condiut area 7.5 ft wide 4 ft deep (ft2) Velocity flow/area (ft/s) R A/P / (P 2w 2d) (ft) Condiut losses L [V/(1.318 V C R0.63)]1/0.54 (ft) where L 412 ft and Hazen-Williams C 120 WSEL at Point 16 (ft) 19. Point 16 to Point 17 Flow Q (ft3/s) Condiut area @ 16 7.5 ft wide 4 ft deep (ft2) Condiut area @ 17 5.5 ft wide 5.5 ft deep (ft2) Average area (ft2) Velocity flow/area (ft/s) R @ 16 A16/ (2 (7.5 ft 4 ft) (ft) R @ 17 A17/ (2 (5.5 ft 5.5 ft) (ft) Average R (ft) Condiut Losses L [V/(1.318 V C R0.63)]1/0.54 (ft) where L 30 ft and Hazen-Williams C 120 WSEL at Point 17 (ft)





20. Point 17 to Point 18 Flow Q (ft3/s) Condiut area @ 17 5.5 ft wide 5.5 ft deep (ft2) Velocity 17 flow/area 17 (ft/s) Pipe area @ 18 5.5 ft2/4 (ft2) Velocity 18 flow/area 18 (ft/s) Exit Losses V182/2g V172/2g (ft) WSEL at Point 18 (ft) 21. Point 18 to Point 19 R A/P / (P d ) (ft) Local losses 3 elbows (3 0.25V 2/2g) entrance (0.5 V 2/2g ) 1.25 V 2/2g (ft) Condiut losses L [V/(1.318 V C R0.63)]1/0.54 where L 455 ft and Hazen-Williams C 120 (ft) WSEL at Point 19 (exit of Control Chamber) (ft)

Design Operation

Min Day

Avg Day

Avg Day Max Hour

77.36 30.25 2.56 23.76 3.26 0.06 360.56

108.31 30.25 3.58 23.76 4.56 0.12 360.96

116.04 30.25 3.84 23.76 4.88 0.14 361.22

154.72 30.25 5.11 23.76 6.51 0.25 362.06

1.38

1.38

1.38

1.38

0.21

0.40

0.46

0.82

0.24 361.00

0.44 361.81

0.50 362.18

0.85 363.74

360.00

360.00

360.00

360.00

1.00 9.00

1.81 9.00

2.18 9.00

3.74 9.00

22. Point 19 to Point 20 Weir elevation (ft) Depth of flow over weir (WSEL @ 19 - weir elevation) (ft) Length of weir, L (ft) Flow over weir q 3.1 h 3/2 x [1–(d/h)3/2]0.385 L Note: Rather than solve for h, find an h by trial and error that gives a q equal to the flow for the given flow scenario (given in Item 1) First Iteration assume h (ft) then q (ft3/s) Second Iteration assume h (ft) then q (ft3/s) Note: These q’s equal the flows for the given scerios (Item 1) h (ft) WSEL at Point 20 (h + WSEL @ Point 19) (ft)

2 66.63 2.17 77.11

3 113.72 2.93 108.41

3 99.77 3.2 116.04

4 90.40 4.5 154.28

2.17 362.17

2.93 362.93

3.2 363.20

4.5 364.50

23. Point 20 to Point 21 Flow Q (ft3/s) Sluice gate area 54 in 54 in (ft2) Velocity = flow/area (ft/s) Gate losses 1.5 V 2/2g (ft) WSEL at Point 21 (Raw Water Control Chamber) (ft)

77.36 20.25 3.82 0.34 362.51

108.31 20.25 5.35 0.67 363.60

116.04 20.25 5.73 0.76 363.96

154.72 20.25 7.64 1.36 365.86

1.76

2.21

2.31

2.80

The overflow weir in the Raw Water Control Chamber is 10 ft long and is sharp crested. Q 3.3 L h 3/2 so h (Q/3.3/L / )2/3 (ft) The water surface must not rise above elevation 370 ft – 0 in. The overflow weir elevation may be safely set at 367 ft – 0 in.



22.96

Chapter Twenty-Two

TABLE 22.6 Hydraulic Calculations in a Medium Sized Water Treatment Plant from the Filter Effluent to the Effluent Clearwell Initial Operation Parameter 1. Plant flow (mgd) (ft3/s)

Design Operation

Min Day

Avg Day

Avg Day Max Hour

50 77.36

70 108.31

75 116.04

100 154.72

345.00 333.00 38.68

345.00 333.00 54.15

345.00 333.00 58.02

345.00 333.00 77.36

120.00 0.64 0.00

120.00 0.90 0.01

120.00 0.97 0.01

120.00 1.29 0.01

120.00 0.64 0.01

120.00 0.90 0.01

120.00 0.97 0.01

120.00 1.29 0.03

0.01 345.02

0.01 345.03

0.01 345.04

0.03 345.06

38.68

54.15

58.02

77.36

23.76 1.63

23.76 2.28

23.76 2.44

0.02 0.02

0.04 0.04

0.05 0.05

23.76 3.26 0.16 0.08 0.08

0.03 345.08

0.05 345.16

0.06 345.19

0.10 345.49

Note: for Points 22 through 28, see Fig. 22.15 2. Point 22 to Point 23 Maximum water level in clearwell (Point 22) (ft) Invert in clearwell, (ft) Flow Q/2 (ft3/s) Stop logs @ A Flow area (2 openings, 5 ft wide, 12 ft deep) (ft2) Velocity flow/area (ft/s) Loss 0.5 V 2/2g (ft) Baffles Flow area (10 ft wide, 12 ft deep) (ft2) Velocity flow/area (ft/s) Loss 1.0 V2/2g (ft) Stop logs @ B and C Same as the losses @ A, times 2 (ft) WSEL at Point 23 (ft) 3. Point 23 to Point 24 Flow Q/2 (ft3/s) 66 inch diameter pipe Flow area d 2/4 p (ft2) Velocity flow/area (ft/s) Exit loss @ clearwell V 2/2g (ft) Loss @ 2 90o bends (0.25 V 2/2g) 2 (ft) Entrance loss @ filter building 0.5 V 2/2g (ft) Pipe loss (3.022 V 1.85 L)/ (C 1.85 D1.165) (ft) where C 120 and L 190 WSEL at Point 24 (ft) 4. Point 24 to Point 25 Flow Q/4 (ft3/s) Flow area 5 ft 5ft (ft2) Velocity Q/A / (ft/s) Loss as flows merge 1.0 V 2/2g (ft) Condiut loss [(V n)/(1.486 R 2/3)] 2 L (ft) where n 0.013, L 55 ft and R A/P / (P 20) WSEL at Point 25 (ft)

19.34 25.00 0.77 0.01

27.08 25.00 1.08 0.02

29.01 25.00 1.16 0.02

38.68 25.00 1.55 0.04

0.00 345.10

0.00 345.19

0.00 345.21

0.01 345.54

5. Point 25 to Point 26 Sluice Gate No. 1 flow area 48 in 36 in (ft2) Velocity Q/A / (ft/s) Loss 0.5 V 2/2g (ft) WSEL at Point 26 (ft)

12 1.61 0.02 345.12

12 2.26 0.04 345.23

12 2.42 0.05 345.26

12 3.22 0.08 345.62

6. Point 26 to Point 27 Sluice Gate No. 2 Loss 0.8 V 2/2g (ft) WSEL at Point 27 (ft)

0.03 345.15

0.06 345.29

0.07 345.33

0.13 345.75





7. Point 27 to Point 28 Port to filter clearwell. Calculate losses through port as if it were a weir when depth of flow is below top of port. Port Dimmensions 9 ft wide by 2 ft – 8 in feet deep Flow Q/4 (ft3/s) Weir (bottom of port) elevation (ft) Depth of flow over weir (WSEL @ 27 weir elevation) (ft) Flow over submerged weir = q = 3.1 h3/2 [1 (d/ d h)3/2]0.385 L Note: Rather than solve for h, find an h by trial and error that gives a q equal to the flow for the given flow scenario (given in item 1). assume h (ft) = then q (ft3s) = assume h (ft) = then q (ft3/s) = Note: These q’s equal the flows for the given scerios (Item 1) h (feet) WSEL at Point 28, (ft)

Design Operation

Min Day

Avg Day

Avg Day Max Hour

19.34 344.00

27.08 344.00

29.01 344.00

38.68 344.00

1.15

1.29

1.33

1.75

1.3 20.8841 1.28 19.4646

1.4 20.2379 1.49 27.0883

1.5 25.5883 1.55 29.232

2 41.0387 1.97 38.485

1.28 345.28

1.49 345.49

1.55 345.55

1.97 345.97

360.00

360.00

360.00

360.00

9.67

13.54

14.51

19.34

0.60

0.85

0.91

1.21

0.00

0.01

0.01

0.02

0.77 0.00 0.00 360.01

1.08 0.00 0.00 360.02

1.15 0.01 0.01 360.02

1.54 0.01 0.01 360.04

8.01 48.05 0.20 2.18

8.02 48.11 0.28 2.18

8.02 48.12 0.30 2.18

8.04 48.22 0.40 2.18

0.00 360.01

0.00 360.02

0.00 360.02

0.00 360.04

Filters–See Filter Hydraulics in Table 22.7 Note: For Points 29 thruogh 33, see Fig. 22.16 8. Point 29 WSEL above filters (ft) 9. Point 29 to Point 30 Entrance to Filter #4 Flow Q/8 (ft3/s) Channel Velocity V Flow/ F /Area (area 4 ft 4 ft) (ft/s) Submerged entrance loss = 0.8 V 2/2g (ft) 48 in Pipe velocity flow/area (area d 2/4 ) (ft/s) Butterfly valve loss 0.25 V 2/2g (ft) Sudden elargement loss 0.25 V 2/2g (ft) WSEL in influent channel (Point 30) (ft) 10. Point 30 to Point 31 Flow depth WSEL @ 30 invert (352 ft) (ft) Flow area 6 ft width depth (ft2) Velocity flow/area (ft/s) R A/P / (P w 2d) (ft) Condiut loss [(V n)/(1.486 R 2/3)] 2 L where n 0.014 and L 35 ft – 4 in (ft) WSEL at Point 31 (ft) 11. Point 31 to Point 32



22.98

Chapter Twenty-Two

TABLE 22.6


Flow Q/4 (ft3/s) Flow depth WSEL @ 31 invert (352 ft) (ft) Flow area 6 ft width depth (ft2) Velocity flow/area (ft/s) R A/P / (P w 2d) (ft) Condiut loss [(V n)/(1.486 R 2/3)] 2 L (ft) where n 0.014 and L 35ft 4 in WSEL at Point 32 (ft) 12. Point 32 to Point 33 Flow 3Q/8 (ft3/s) Flow depth WSEL @ 32 invert (352 ft) (ft) Flow area 6 – ft width depth (ft2) Velocity flow/area (ft/s) R A/P / (P w 2d) (ft) Condiut loss [(V n)/(1.486 R 2/3)] 2 L (ft) where n 0.014 and L 35 ft – 4 in WSEL at Point 33 (ft) 13. Point 33 to Point 1 Flow Q/2 (ft3/s) Flow depth WSEL @ 33 invert (352 ft) (ft) Flow area 6 ft width depth (ft3) V Velocity flow/area (ft/s) R A/P / (P w 2d) (ft) Condiut loss [(V n)/1.486 R 2/3)] 2 L (ft) where n 0.014 and L 36 ft – 4 in WSEL at Point 1 (ft)

TABLE 22.7

Design Operation

Min Day

Avg Day

Avg Day Max Hour

19.34 8.01 48.06 0.40 2.18

27.08 8.02 48.11 0.56 2.18

29.01 8.02 48.12 0.60 2.18

38.68 8.04 48.22 0.80 2.18

0.00 360.01

0.00 360.02

0.00 360.02

0.00 360.04

29.01 8.01 48.06 0.60 2.18

40.61 8.02 48.11 0.84 2.18

43.52 8.02 48.13 0.90 2.18

58.02 8.04 48.22 1.20 2.18

0.00 360.01

0.00 360.02

0.00 360.02

0.00 360.04

38.68 8.01 48.06 0.80 2.18

54.15 8.02 48.11 1.13 2.18

58.02 8.02 48.13 1.21 2.18

77.36 8.04 48.23 1.60 2.18

0.00 360.01

0.00 360.02

0.00 360.02

0.00 360.04

Example Hydraulic Calculation of a Typical Filter Initial Operation Parameter

Plant Flow (mgd) Filter loading, gpm/ft2 Filter area per filter – 7 out of 8 Filters in Operation (ft2) Flow = loading area (gal/min) (mgd) (ft3/s) Losses through filter effluent piping (Fig. 22.17) 20 in piping (Q): Pipe velocity Q/A / (ft/s) Local losses Exit (0.5) butterfly valves (2 0.25) + 90o Elbows (2 0.4) tee (1.8) 3.6 V 2/2g (ft) R A/P / (d 2/4 )/(d ) dd/4 (ft) Condiut losses L [V/(1.318 V C R 0.63)]1/0.54 where L 20 ft and Hazen-Williams C 120 (ft) 20 in piping (Q/2):

Design Operation

Min Day

Avg Day

Avg Day Max Hour

50 2 1240

70 4 1240

75 6 1240

100 8 1240

2480 3.57 5.53

4960 7.14 11.05

7440 10.71 16.58

9920 14.29 22.10

2.53

5.07

7.60

10.13

0.36 0.42

1.43 0.42

3.23 0.42

5.74 0.42

0.03

0.09

0.20

0.34





Min Day

Pipe velocity Q/A (ft/s) Local losses butterfly valve (0.25) (ft) R A/P / (d 2/4 )/(d ) dd/4 (ft) Condiut losses L [V/(1.318 V C R 0.63)]1/0.54 where L 10 ft and Hazen-Williams C 120 (ft) 24 in piping (Q/2): Pipe velocity Q/A, (ft/s) Local losses entrance (1.0) Tee (1.8) 2.8 V 2/2g (ft) Filter (clean) and underdrain losses (obtain from manufacturer) (ft) Total losses (effluent pipe and clean filters) (ft)

Avg Day


1.27 0.01 0.42

2.53 0.02 0.42

3.80 0.06 0.42

5.07 0.10 0.42

0.00

0.01

0.03

0.05

0.88

1.76

2.64

3.52

0.03

0.13

0.30

0.54

0.30 0.73

0.50 2.20

0.75 4.57

1.10 7.87

Assume that headloss will be allowed to incrase eight ft before the filters are backwashed. A rate controller will be used to maintain a constant flow through the filter. Determine the ranges of available head over which the rate controller will operate. Static head see figure 2.18 WSEL above filters (ft) WSEL in filter effluent conduit, Point 29 (see Example 22–2) Maximum (ft) Minimum (ft) Static head WSEL above filters – WSEL in filter effluent condiut Maximum (ft) Minimum (ft) Available head static head 8 ft Maximum (ft) Minimum (ft)

TABLE 22.8

360.00

360.00

360.00

360.00

346.50 345.00

346.50 345.00

346.50 345.00

346.50 345.00

15.00 13.50

15.00 13.50

15.00 13.50

15.00 13.50

7.00 5.50

7.00 5.50

7.00 5.50

7.00 5.50

Example Hydraulic Calculation of a Typical Bar Screen System Initial Operation Parameter

1. Wastewater flow rate (ft3/s) (mgd) Bar screens Total of number of units Number of units in operation Number of units in standby Flow rate per screen in operation, q (ft3/s) Width of each bar screen, w (ft) 2. At Point 8 Pump wetwell HGL at high water level, HGL7 (ft)

Min Day Avg Day

Design Operation Avg Day Max Hour Max Hour

35.3 23

56.5 37

70.6 46

113.0 73

113.0 73

3 2 1 17.1 8.2

3 2 1 28.3 8.2

3 2 1 35.3 8.2

3 3 0 37.7 8.2

3 3 1 56.5 8.2

330.05

330.05

330.05

330.05

330.05



22.100

Chapter Twenty-Two

TABLE 22.8


Min Day Avg Day

(pump starts at EL 330.05 ft and stops at EL 328.08 (ft)) Pump well bottom EL (ft) Check for critical depth in bar screen channel: Critical depth in a rectangular channel Yc (q2/g / /w2)(1/3) (ft) Bar screen channel depth pump WW HGL channel bottom ELl (ft) (water level at pump well controls upstream hydraulics if normal depth is higher than Yc) Io bar screen channel depth higher than yc 3. Point 8 to Point 7 Channel bottom EL (ft) Depth in channel, y7 (ft) Velocity, V V7 (ft/s) Exit loss from channel to pump well Exit loss coefficient Kexit 1.0 Headloss Kexit V7 V 2/2g, Hle7 (ft) HGL at Point 7, HGL7= HGL8 le(ft) 4. Point 7 to Point 6 Friction headloss through channel Length of approach channel, L6 (ft) Manning's number n for concrete channel Channel width, w6 (ft) Water depth, h6 (ft) Velocity, V V6 (fps) Hydraulic radius, R6 (h6 w6)/(2 h6w6) (ft) Headloss (V6 V xn/1.486 R6 (2/3))2 L6, Hlf6 f (ft) HGL at Point 6, HGL6 HGL7 Hlf6 f (ft) 5. Point 6 to Point 5 Calculate headloss through bar screen Space between bars (ft) Bar width (ft) Bar shape factor, bsf Cross sectional width of bars, w (ft) Clear spacing of bars, b (ft) Upstream velocity head, h (ft) Angle of bar screen with horizontal, p (degrees) Kirschmer’s eq. Hls bsf w/b 1.33 h sin p (ft) Allow 6 in head for blinding by screenings, Ha (ft)


324.80

324.80

324.80

324.80

324.80

0.52 3.61

0.72 3.61

0.83 3.61

0.87 3.61

1.14 3.61

yes

yes

yes

yes

yes

326.44 3.61 0.60

326.44 3.61 0.95

326.44 3.61 1.19

326.44 3.61 1.27

326.44 3.61 1.91

1.0 0.01

1.0 0.01

1.0 0.02

1.0 0.03

1.0 0.06

330.06

330.07

330.07

330.08

330.11

23 0.013 8.20 3.61 0.60

23 0.013 8.20 3.62 0.95

23 0.013 8.20 3.63 1.19

23 0.013 8.20 3.63 1.26

23 0.013 8.20 3.67 1.88

1.92

1.92

1.93

1.93

1.94

0.00

0.00

0.00

0.00

0.00

330.06

330.07

330.08

330.08

330.11

0.06 0.033 2.42 2.93 5.27 0.0134

0.06 0.033 2.42 2.93 5.27 0.0342

0.06 0.033 2.42 2.93 5.27 0.0535

0.06 0.033 2.42 2.93 5.27 0.0608

0.06 0.033 2.42 2.93 5.27 0.1369

60

60

60

60

60

0.02

0.05

0.08

0.09

0.21

0.5

0.5

0.5

0.5

0.5





HGL upstream of bar screen, HGL5 HGL6 HlsHa (ft)

Min Day Avg Day


330.58

330.62

330.66

330.67

330.82

6. Point 5 to Point 4 Friction headloss through channel Length of approach channel, L4 (ft) 22.97 Manning’s n for concrete channel 0.013 Channel width, w4 (ft) 8.20 Channel bottom elevation (ft) 326.94 Water depth, h4 (ft) 3.64 Channel velocity, V V4 (ft/s) 0.59 Hydraulic radius R4h4w4/(2h4w4) 1.93 Headloss (V4 V n/1.486 R4(2/3) )2 L4, Hlf4 f (ft) 0.00

22.97 0.013 8.20 326.94 3.68 0.93 1.94

22.97 0.013 8.20 326.94 3.72 1.16 1.95

22.97 0.013 8.20 326.94 3.74 1.23 1.96

22.97 0.013 8.20 326.94 3.89 1.77 2.00

0.00

0.00

0.00

0.00

330.58

330.62

330.66

330.67

330.83

1.0 3.94 2.95 1.23

1.0 3.94 2.95 1.95

1.0 3.94 2.95 2.41

1.0 3.94 2.95 2.56

1.0 3.94 2.95 3.69

HGL at Point 4, HGL4 HGL5 Hlf4, f ft 7. Point 4 to Point 3 Headloss at sluice gate contraction Kgate Sluice gate width (ft) Sluice gate heigth (ft) Velocity through sluice gate, Vs (ft/s) Sluice gate headloss, Hls Kgate Vs 2/2g (ft) HGL at Point 3, HGL3 (ft) 8. Point 3 to Point 2 Water depth at Point 2, h2 (ft) Channel width, w2 (ft) Channel velocity, V V2 (ft/s) Fitting headloss through a 45° bend, Kbend 0.20 V 2/2g, Hlb2 (ft) Headloss Kbend V2 Friction headloss through channel Length of approach channel, L2 (ft) Manning’s n for concrete channel Hydraulic radius R2 h2 w2/ (2 f 2 w2) (ft) Headloss (V n/1.486 R2(2/3))2 L2, Hlf2 f (ft) Entrance loss Kent 0.5 2 Headloss Kent V 2 /2g, Hle2 (ft) HGL at point 2, HGL2 HGL3 Hlb2 Hlf2 f Hle2 (ft) 9. Point 2 to Point 1 HGL at Point 1, HGL 1 HGL2 (ft) Invert EL of inlet sewer, INV1 (ft) Crown EL of inlet sewer, CWN1 (ft) Surcharge to inlet sewer?

0.02

0.06

0.09

0.10

0.21

330.60

330.68

330.75

330.75

331.04

3.67 6.56 0.73

3.74 6.56 1.15

3.81 6.56 1.41

3.84 6.56 1.49

4.10 6.56 2.10

0.20 0.0017

0.20 0.0041

0.20 0.0062

0.20 0.0069

0.0137

13.12 0.013

13.12 0.013

13.12 0.013

13.12 0.013

13.12 0.013

1.73

1.75

1.76

1.77

1.82

0.00

0.00

0.00

0.00

0.00

0.50 0.0042

0.50 0.0103

0.50 0.0155

0.50 0.0174

0.50 0.0342

330.61

330.69

330.77

330.80

331.09

330.61 326.44 333.50 No

330.69 326.44 333.50 No

330.77 326.44 333.50 No

330.80 326.44 333.50 No

331.09 326.44 333.50 No



22.102

Chapter Twenty-Two

TABLE 22.9

Example Hydraulic Calculation of a Typical Vortex Grit Tank System Initial Operation Parameter

1. Wastewater flow rate, Q (cfs) (mgd) 2. Vortex grit tanks total number of units Number of units in operation Number of units in standby Flow rate per vortex grit tank in operation, q (cfs)

Min Day Avg Day


Peak

35.3 23

56.5 36

70.6 46

113.0 73

113.0 73

3 2 1

3 2 1

3 2 1

3 3 0

3 2 1

17.7

28.3

35.3

37.7

56.5

347.79 344.51 17.7 9.84

347.79 344.51 28.3 9.84

347.79 344.51 35.3 9.84

347.79 344.51 37.7 9.84

347.79 344.50 56.5 9.84

0.67 0.67 0.00

0.92 0.92 0.00

1.06 1.06 0.00

1.11 1.11 0.00

1.46 1.46 0.00

348.45

348.70

348.85

348.90

349.25

9.84 344.49 3.96 0.45

9.84 344.49 4.21 0.68

9.84 344.49 4.36 0.82

9.84 344.49 4.41 0.87

9.84 344.49 4.46 1.21

1.0 0.0032

1.0 0.0072

1.0 0.0105

1.0 0.0117

1.0 0.0226

348.46

348.71

348.86

348.91

349.27

8.20 344.49 3.97 0.54

8.20 344.49 4.22 0.82

8.20 344.49 4.37 0.98

8.20 344.49 4.42 1.04

8.20 344.49 4.78 1.44

32.81 0.013

32.81 0.013

32.81 0.013

32.81 0.013

32.81 0.013

2.02

2.08

2.12

2.13

2.21

0.0003

0.0006

0.0009

0.0010

0.0018

1.0

1.0

1.0

1.0

1.0

Control point is located at channel weir Hydraulic Calculations Upstream of Control point 3. At Point 8 Headloss over sharp-crested weir Sharp-crested weir EL, weir EL (ft) Effluent channel bottom EL (ft) Flow rate over weir, q (ft3/s) Length of weir, L (ft) Head over end contracted weir, He (assumed) [q/ 33.3(L–0.2 L He)](2/3) (ft) Hle8 – He (must be zero) HGL at Point 8, HGL8 weir EL Hle8 (ft) 4. Point 8 to Point 7 Channel width, w7 (ft) Channel bottom EL (ft) Water depth, h7 (ft) Velocity, V V7 (ft/s) Exit headloss from channel to effluent weir Exit headloss coefficient Kexit 1.0 2 Headloss, Hle7 Kexit V 7 /2g (ft) HGL at Point 7, HGL7 HGL8 Hle7 (ft) 5. Point 7 to Point 6 Channel width, w6 (ft) Channel bottom EL (ft) Water depth, h6 (ft) Velocity, V V6 (ft/s) Friction headloss through channel Length of approach channel, L6 (ft) Manning’s n for concrete channel Hydraulic radius, R6 (h6 w6)/ (2 h6 w6) (ft) 2 Headloss (V6 V n/R / 6(2/3)) L6, Hlf6 f (ft) Fitting headloss through 90° bend Fitting headloss coefficient Kbend 1.0





Min Day Avg Day

Headloss Kbend V6 V /2g, Hlb6 (ft) 2

HGL at Point 6, HGL6 HGL7 Hlf6 f Hlb6 (ft) 6. Point 6 to Point 5 Headloss through sluice gate Sluice gate headloss coefficient Kgate 1.0 Sluice gate width (ft) Sluice gate height (ft) Water depth, h5 (ft) Sluice gate height or h5, whichever smaller Velocity through sluice gate, V V5 (ft/s) 2 Headloss, Hls5 Kgate V5 /2g (ft) HGL at Point 5, HGL5 HGL6 Hls5 (ft) 7. Point 5 to Point 4 Channel width, w4 (ft) Bottom of channel EL (ft) Water depth, h4 (ft) Channel velocity, V V4 (ft/s) Fitting headloss through a 90° bend Fitting headloss coefficient Kbend 1.0 Headloss, Hlb4 Kbend V4 V 2/2g (ft) Friction headloss through channel Length of approach channel, L4, (ft) Manning’s n for concrete channel Hydraulic radius R4 h4 w4/ (2 h4 w4) (ft) Headloss, Hlf4 f (V4 V n/1.486 R4(2/3))2 L4 (ft) HGL at Point 4, HGL4 HGL5 Hlb4 Hlf4 f (ft) 8. Point 4 to Point 3 Headloss across vortex grit tank, H1tank (ft) (per manufacturer recommendations) HGL at Point 3, HGL3 HGL4 H1tank (ft) 9. Point 3 to Point 2 Channel width, w2, (ft) Bottom of channel EL (ft) Water depth, h2 (ft) Channel velocity, V V2 (ft/s) Friction headloss through channel Length of approach channel, L2 (ft) Manning’s n for concrete channel Hydraulic radius R2 h2R4(2/3))2 w2/(2*h2 w2) (ft)


Peak

0.0046

0.0103

0.0151

0.0168

0.0322

348.46

348.72

348.88

348.93

349.31

1.0 4.92 3.28 3.97

1.0 4.92 3.28 4.22

1.0 4.92 3.28 4.37

1.0 4.92 3.28 4.42

1.0 4.92 3.28 4.78

3.28 1.09 0.0186

3.28 1.75 0.0475

3.28 2.19 0.0743

3.28 2.33 0.0845

3.28 3.50 0.1902

348.48

348.77

348.95

349.01

349.50

8.20 345.14 3.34 0.65

8.20 345.14 3.62 0.95

8.20 345.14 3.81 1.13

8.20 345.14 3.87 1.19

8.20 345.14 4.35 1.58

1.0 0.0065

1.0 0.0140

1.0 0.0199

1.0 0.0219

1.0 0.0389

32.81 0.013

32.81 0.013

32.81 0.013

32.81 0.013

32.81 0.013

1.84

1.92

1.97

1.99

2.11

0.0005

0.0009

0.0013

0.0014

0.0023

348.49

348.78

348.97

349.04

349.54

0.20

0.20

0.20

0.20

0.20

348.68

348.98

349.17

349.23

349.73

6.56 346.46 2.23 1.21

6.56 346.46 2.52 1.71

6.56 346.46 2.71 1.98

6.56 346.46 2.78 2.07

6.56 346.46 3.28 2.63

45.93 0.013

45.93 0.013

45.93 0.013

45.93 0.013

45.93 0.013

1.33

1.43

1.48

1.50

1.64



22.104

Chapter Twenty-Two

TABLE 22.9


Min Day Avg Day

Headloss, Hlf2 f (V2* V n/1.486 R ) L2 (ft) 0.0035 Headloss through sluice gate Sluice gate headloss coefficient Kgate 1.0 1.0 Sluice gate width (ft) 4.9 Sluice gate height (ft) 3.3 Water depth, h2 (ft) 2.23 Sluice gate height or h2, whichever smaller (ft) 2.23 Velocity through sluice gate (ft/s) 1.61 2 Headloss, Hls2 Kgate V2 V /2g (ft) 0.0403


Peak

(2/3) 2

HGL at point 2, HGL2 HGL3 Hlf2 f Hls2 (ft) 10. Point 2 to Point 1 Channel width, w1 (ft) Bottom of channel EL (ft) Water depth, h1 (ft) Channel velocity, V1 (ft/s) Fitting headloss through a 90° deg bend Fitting headloss coefficient Kbend 1.0 Headloss, Hlb1 Kbend V12/2g (ft) Friction headloss through channel Length of approach channel, L1 (ft) Manning’s n for concrete channel Hydraulic radius R1 h1 w1/(2 h1 w1) (ft) Headloss, Hlf1 f (V1 n/1.486 R1(2/3) )2 L1 (ft)

0.0064

0.0082

0.0087

0.0125

1..0 4.9 3.3 2.52 2.52 2.28 0.0804

1.0 4.9 3.3 2.71 2.71 2.65 0.1087

1.0 4.9 3.3 2.78 2.78 2.76 0.1181

1.0 4.9 3.3 3.28 3.28 3.50 0.1905

348.73

349.07

349.29

349.36

349.94

6.56 346.62 2.11 1.28

6.56 346.62 2.44 1.76

6.56 346.62 2.67 2.02

6.56 346.62 2.74 2.10

6.56 346.62 3.32 2.60

1.0 0.0253

1.0 0.0482

1.0 0.0633

1.0 0.0683

1.0 0.1046

16.40 0.013

16.40 0.013

16.40 0.013

16.40 0.013

16.40 0.013

1.28

1.40

1.47

1.49

1.65

0.0015

0.0025

0.0031

0.0032

0.0043

348.75

349.12

349.35

349.43

350.05

(Influent channel may be aerated using diffused air to prevent solids settling or odor problem) HGL at Point 1, HGL1 HGL2 Hlb1 Hlf1 f (ft)

TABLE 22.10 Example Hydraulic Calculation of a Typical Primary Sedimentation Tank System Initial Operation Design Operation Parameter 1. Wastewater flow rate, Q (ft3/s) (mgd) 2. Primary sedimentation tanks (PSTs) Total number of units Number of units in operation Number of units on standby Flow rate per PTS in operation, q (ft3/s)

Min Day Avg Day

Avg Day Max Hour

Peak

35.3 23

56.5 37

70.6 46

113.0 73

113.0 73

3 2 1 17.7

3 2 1 28.3

3 3 0 23.5

3 3 0 37.7

3 2 1 56.5





Min Day Avg Day


Peak

Control points are located at Points 5 and 6 so that back up from down stream does not flood effluent channel or overflow weir. Hydraulic Calculations beginning at Point 7 3. At Point 7 HGL7 must be equal to HGL1 of aeration tank (ft) 4. At Point 6 Allowance of 0.33 ft from HGL at pipe entrance to bottom of PST effluent trough at discharge end (ft) Elevation of PTS trough bottom at discharge end, EL dcb (ft)

342.73

342.73

342.73

342.73

342.73

0.33

0.33

0.33

0.33

0.33

343.06

343.06

343.06

343.06

343.06

147.6

Calculation of water depth in PST effluent trough Tank diameter, Dt (ft)

147.6

147.6

147.6

147.6

Number of channels per tank nc

2

2

2

2

2

Total flow through tank, q (ft3/s)

17.66

28.25

23.54

37.67

56.50

8.83

14.13

11.77

18.83

28.25

Flow per channel, qc q/nc (ft3/s) Channel slope, Sc (selected to prevent solids setting) Channel width, w6 (ft)

0.20

0.20

0.20

0.20

0.20

3.28

3.28

3.28

3.28

3.28 229.23

Channel length, Lc 3.14 229.23

229.23

229.23

229.23

Change in channel EL, ELdif Sc Lc (ft)

(Dt-(w6/2))/nc (ft)

0.46

0.46

0.46

0.46

0.46

Critical depth, yc (qc2/(g w62))0.33 (ft)

0.62

0.84

0.75

1.02

1.33

Water depth at upstream end of channel, yu

0.69

1.07

0.91

1.38

1.92

343.52

343.52

343.52

343.52

343.52

343.68

343.91

343.81

344.08

344.39

344.21

344.59

344.43

344.90

345.44

0.33

0.33

0.33

.033

0.33

[2 (yc)2 (yc (S L/3) L 2 ]0.5 (2 S L L/3) (ft) Channel bottom El at upstream end of trough, ELucb ELdcb ELdif (ft) HGL at trough downstream, HGL6d ELdcb yc (ft) HGL at trough upstream, HGL6u ELucb yu (ft) 5. Point 6 to Point 5 Allowance to Weir from high trough HGL (ft)



22.106

Chapter Twenty-Two

TABLE 22.10


Weir elevation, Elwe, max. HGL6u allowance (ft) Headloss over V–notch weirs Number of weirs per tank, Nw Tank diameter, Dt (ft) Weir length, Lw (Dt) 3.14 (ft) Hydraulic load, So q/Lw / ,(ft3/s/s/ft) Weir angle, A (°) V–notch height, Vh (ft) V–notch width, Vw 2 (TAN(A ( /2)) Vh (ft) Space between notches, Esv (ft) Number of notches per weir, nv Lw/(Ew Esv) Flow per notch, Qcw q/nv Weir coefficient for 90° notch, Cw Water depth over the weir, hle5 (Qcw/Cw)(1/2.48) (ft) hle5 Vh? (if not, need to readjust calculations) HGL at point 5, HGL5 ELwe hle5, (ft) 6. Point 5 to Point 4 Headloss through primary sedimentation tanks Number of tanks, Nt Flow per tank, q (ft3/s) Tank diameter, Dt (ft) Side water depth, Dsw (ft) Tank bottom elevation, ELt HGL5 Dsw (ft) Tank floor slope, St (%) Minimum floor tank elevation, Eltf 0.0833 (Dt/2) t ELt (ft) Headloss through tank, hlt4 t (ft) (Available from equipment manufacturer) HGL at point 4, HGL4 HGL5 hlt4 t (ft) 7. Point 4 to Point3 Headloss through PST influent pier Pier diameter, Dp 42 in Pier length, Lp (ft) Velocity, V3 V Qt/(3.14 t (Dp/2)2 ) (ft/s) Hazen-Williams coefficient, Cp Hydraulic radius, Rp Dp/4 (ft)

Min Day Avg Day


Peak

345.77

345.77

345.77

345.77

345.77

1 147.64 463.58 0.0381 90.00 0.33

1 147.64 463.58 0.0609 90.00 0.33

1 147.64 463.58 0.0508 90.00 0.33

1 147.64 463.58 0.0813 90.00 0.33

1 147.64 463.58 0.1219 90.00 0.33

0.66 0.10

0.66 0.10

0.66 0.10

0.66 0.10

0.66 0.10

614 0.0288 2.43 0.17

614 0.0460 2.43 0.20

614 0.0383 2.43 0.19

614 0.0614 2.43 0.23

614 0.0920 2.43 0.27

Yes

Yes

Yes

Yes

Yes

345.93

345.97

345.95

345.99

346.03

2 17.66 147.64 14.11

2 28.25 147.64 14.11

3 23.54 147.64 14.11

3 37.67 147.64 14.11

2 56.50 147.64 14.11

331.85 8.33 325.70

331.85 8.33 325.70

331.85 8.33 325.70

331.85 8.33 325.70

331.85 8.33 325.70

0.16

0.16

0.16

0.16

0.16

346.10

346.13

346.12

346.16

346.20

3.51 21.33 1.83 120 0.88

3.51 21.33 2.92 120 0.88

3.51 21.33 2.43 120 0.88

3.51 21.33 3.89 120 0.88

3.51 21.33 5.84 120 0.88





Min Day Avg Day


Peak

Slope, Sp [V3/(1.318 V Cp Rp(0.63))](1/0.54) (%) Headloss, Hlf3 f Lp Sp (ft) Exit headloss from pier Exit headloss coefficient Kexit 1.0 2 Headloss, hle3 K V3 V /2g (ft)

0.03 0.0064

0.07 0.0153

0.05 0.0109

0.12 0.0261

0.26 0.0552

1 0.0517

1 0.1324

1 0.0920

1 0.2354

1 0.5297

HGL at Point 3, HGL3 HGL4 Hlf3 f hle3 (ft)

346.16

346.28

346.22

346.42

346.78

3

3

3

3

3

1 3.94 17.66 1.45

1 3.94 28.25 2.32

1 3.94 23.54 1.93

1 3.94 37.67 3.10

1 3.94 56.50 4.64

120 0.98 229.7

120 0.98 229.7

120 0.98 229.7

120 0.98 229.7

120 0.98 229.7

0.02 0.0395

0.04 0.0942

0.03 0.0672

0.07 0.1605

0.15 0.3402

0.05 0.0164

0.05 0.0419

0.05 0.0291

0.05 0.0744

0.05 0.1674

346.21

346.42

346.32

346.65

347.29

3.94 17.66 1.45

3.94 28.25 2.32

3.94 23.54 1.93

3.94 37.67 3.10

3.94 56.50 4.64

0.50 0.0164

0.50 0.0419

0.50 0.0291

0.50 0.0744

0.50 0.1674

346.23

346.46

346.35

346.73

347.46

0.33

0.33

0.33

0.33

0.33

347.79

347.79

347.79

347.79

347.79

8. Point 3 to Point 2 Total number of pipes Number of pipes per primary sedimentation tank Pipe diameter, Dp (ft) Flow per pipe, q (cfs) Velocity, V V2 Friction headloss through primary sedimentation tank influent pipe Hazen-Williams coefficient, Cp Hydraulic radius, Rp Dp/4 (ft) Length of pipe, Lp (ft) Slope, Sp [V2/(1.318 V Cp (1/0.54) (%) Rp (0.63))] Headloss, hlf2 f Lp Sp (ft) Fitting headloss through two 45° bends Fitting headloss coefficient Kbend 0.5 Headloss, hlb2 K V2 V 2/2g (ft) HGL at Point 2, HGL2 HGL3 hlb2 hlf2 f (ft) 9. At Point 1 Entrance headloss from primary sedimentation tank influent distribution box to influent pipe Pipe diameter, Dp (ft) Flow per pipe, q (ft3/s) Velocity, V1 (ft3/s) Entrance headloss coefficient Kentrance 0.5 Headloss, Hle1 Kentrance V12/2g (ft) HGL at point 1, HGL1 HGL2 Hle1 (ft) Allowance to grit tank effluent weir from maximum HGL1, Hall (ft) Grit tank effluent elevation, ELgr HGL1 Hall (ft)



22.108

Chapter Twenty-Two

TABLE 22.11

Example Hydraulic Calculation of a Typical Final Sedimentation Tank Initial Operation Parameter

1. Wastewater flow rate, Q (ft3/s) (mdg) RAS flow, % of average day flow RAS flow, Qras Q RAS flow/100 (ft3/s) Final sedimentation tank influent flow, Qin (ft3/s)

Min Day Avg Day


Peak

35.31 23 20 11.30

56.50 37 50 28.25

70.63 46 50 35.31

113.01 73 100 70.63

113.01 73 100 70.63

46.62

84.76

105.94

183.64

183.64

Final sedimentation tank effluent flow, Qeff (ft3/s)

35.31

56.50

70.63

113.01

113.01

Final sedimentation tanks Total number of units Number of units in operation Number of units on standby Tank width (ft) Influent per operating tank, qin (ft3/s) Effluent per operating tank, qeff (ft3/s)

4 3 1 52 15.54 11.77

4 3 1 52 28.25 18.83

4 3 1 52 35.31 23.54

4 4 0 52 45.91 28.25

4 3 1 52 61.21 37.67

20 23.0 459.3 90.0 0.33

20 23.0 459.3 90.0 0.33

20 23.0 459.3 90.0 0.33

20 23.0 459.3 90.0 0.33

20 23.0 459.3 90.0 0.33

0.66 0.10

0.66 0.10

0.66 0.10

0.66 0.10

0.66 0.10

608 0.0194 2.43

608 0.0310 2.43

608 0.0387 2.43

608 0.0465 2.43

608 0.0620 2.43

0.14

0.17

0.19

0.20

0.23

Yes

Yes

Yes

Yes

Yes

339.16

339.16

339.16

339.16

339.16

339.30

339.33

339.35

339.36

339.38

0.00

0.00

0.00

0.00

0.00

339.30

339.33

339.35

339.36

339.38

2. Select control Point at Point 3 (where effluent wiers are located) Hydraulic calculations downstream of control point At Point 3 V-notch weir Number per tank, Nw Individual weir length, Lw (ft) Total weir length, Lwt Lw Nw (ft) Weir angle, A° V-notch height, Vh (ft) V-notch width, Vw 2 (TAN(A ( /2)) Vh (ft) Space between notches, Esv (ft) Total number of notches per tank, nv Lwt/( t Vw Esv) Flow per notch, Qcw qeff /nv Weir coefficient for 90° notch, Cw Water depth over the weir, hle3 (Qcw/Cw)(1/2.48) (ft) hle3 Vh? (if not, need to readjust calculations) Weir EL (ft) (select weir elevation so that HGL1 equals aeration tank’s HGL6) HGL at Point 3, EGL3 Weir EL hle3 (ft) Velocity head, HV 0 (assume V3 V 0) (ft) HGL at point 3, HGL3 weir EL hle3 (ft)





Min Day Avg Day

3. Point 3 to Point 4 Effluent troughs Number of troughs, nt 10 Flow per trough, qt qeff /nt 1.18 Trough slope, St (%) (select to prevent solids settling) 0.20 Trough width, w6 (ft) 1.6 Approximate trough length, Lt (ft) 23.0 Change in trough EL, difEL4 St Lt (ft) 0.05 Critical depth, yc (qt 2/(gw62)0.33 (ft) 0.26 Water depth at upstream end of trough for free fall 0.41 from trough into final effluent channel yu4 [2 (yc)2 (yc-(S L/3)) L 2]0.5 (2 S L L/3) (ft) Max water EL downstream of weir (occuring at max, hour flow with one tank out of service), Elmax4 weir EL-0.33 ft (see Point 3 for weirEL) Trough bottom EL at upstream end of trough, TbuEL4 ft 337.90 TbuEL4 ELmax4 yu for max hour flow with one tank out of service HGL at upstream end, HGL4u TbuEL4 yu4 (ft) Velocity head, HV4 V u0 (assume V 0) (ft) EGL at upstream end, EGL4u HGL4u HV4 V u (ft) Trough bottom EL at downstream end of trough Tbd EL4 TbuEL4 dif EL4 (ft) HGL at Point 4, HGL4 TbdEL4 yc (ft) Velocity head, HV4 V d = Vc2/ 2g (ft) EGL at upstream end, EGL4u HGL4u HV4 V u, ft 4. Point 4 to Point 5 Effluent channel upstream Max. water surface level at upstream end of effluent channel, ELmax5 = TbdEL4-0.33 (ft) HGL at Point 5, HGL5 ELmax5 (ft) Velocity head, HV5 V 0 (assume V 0) (ft) EGL maximum at point 5, EGL5m HGL5m HV5 V (ft) 5. Point 5 to Point 6 Effluent channel downstream Flow through channel, Qeff (ft3/s)


Peak

10 1.88

10 2.35

10 2.83

10 3.77

0.20 1.6 23.0 0.05 0.35

0.20 1.6 23.0 0.05 0.41

0.20 1.6 23.0 0.05 0.46

0.20 1.6 23.0 0.05 0.56

0.57

0.67

0.76

0.93

338.83

337.90

337.90

337.90

337.90

338.31

338.47

338.57

338.66

338.83

0.00

0.00

0.00

0.00

0.00

338.31

338.47

338.57

338.66

338.83

337.86

337.86

337.86

337.86

337.86

338.12 0.39

338.21 0.54

338.27 0.63

338.32 0.71

338.41 0.87

338.51

338.75

338.90

339.03

339.28

337.53

337.53

337.53

337.53

337.53

337.53 0.00

337.53 0.00

337.53 0.00

337.53 0.00

337.53 0.00

337.53

337.53

337.53

337.53

337.53

35.31

56.50

70.63

113.01

113.01



22.110

Chapter Twenty-Two

TABLE 22.11


Min Day Avg Day

Channel slope, Sc (%) (select to prevent solids settling) 0.20 Channel width, w6 (ft) 9.8 Approximate channel length, Lch (ft) 210.0 Change in channel EL, difEL6 Sc Lch (ft) 0.42 Critical depth, yc (q2/(gw62)0.33 (ft) 0.75 Water depth at upstream end of channel, 0.94 yu6 [2 (yc)2 (yc (S L/3))2]0.5 (2 S L/3), ft Channel bottom EL at upstream end of channel, 335.10 cbuEL6 HGL5- maximum yu6 (ft) HGL at upstream end of channel, HGL5 cbuEL6 yu6 (ft) Velocity head, HV5 V 0 (assume V 0) (ft) EGL at upstream end of channel, EGL5 HGL5 HV5 V (ft) Channel bottom EL at downstream end of channel, cbdEL6 cbuEL6 difEL6 (ft) HGL at Point 6, HGL6 cbdEL6 yc (ft) Velocity head, HV6 V Vc2/2g (ft) EGL at Point 6, EGL6 HGL6 HV6 V (ft) 6. At Point 7 Max. water EL downstream of channel end free-fall HGL at Point 7, HGL7 cbdEL6 0.33 (ft) (This must be the same as maximum elevation at Point 1 of multi-media filter.)


Peak

0.20 9.8 210.0 0.42 1.02 1.41

0.20 9.8 210.0 0.42 1.18 1.69

0.20 9.8 210.0 0.42 1.61 2.43

0.20 9.8 210.0 0.42 1.61 2.43

335.10

335.10

335.10

335.10

336.04

336.51

336.79

337.53

337.53

0.00

0.00

0.00

0.00

0.00

336.04

336.51

336.79

337.53

337.53

334.68

334.68

334.68

334.68

334.68

335.42 1.17 336.60

335.70 1.62 337.31

335.86 1.88 337.74

336.29 2.59 338.88

336.29 2.59 338.88

334.35

334.35

334.35

334.35

334.35

3

3

3

4

3

15.54 52.5 393.7 325.4 13.92

28.25 52.5 393.7 325.4 13.95

35.31 52.5 393.7 325.4 13.97

45.91 52.5 393.7 325.4 13.98

61.21 52.5 393.7 325.4 14.01

Hydraulic calculations upstream of control point 7. At Point 2 Final sedimentation tanks (Gould type) Number of tanks in operation, nt Flow per tank upstream of sludge collection, qin (ft3/s) Tank width, Wt (ft) Tank length, Lt (ft) Tank bottom elevation at influent end (ft) Side water depth (ft) Assume friction losses, Hlf2, f through tank are negligible EGL at Point 2, EGL2 EGL3 Hlf2 f (ft) Velocity head, HV2 V 0 (assume V 0) (ft) HGL at Point 2, HGL2 EGL3 HV2 V (ft)

0.0

0.0

0.0

0.0

0.0

339.30 0.00 339.30

339.33 0.00 339.33

339.35 0.00 339.35

333.36 0.00 333.36

339.38 0.00 339.38

8. Point 2 to Point 1





Tank influent sluice gates Height (ft) Width, Ws (ft) Area (ft2) Number of sluice gates per tank, Nsg Flow per sluice gate, qsg qin /Nsg / , (ft3/s) Upstream head over weir, Du (select so Qsub qsg 0) (ft) Downstream head over weir, Dd (qsg/3.33/Ws')(2/3) (ft) Effective sluice gate width, Ws' Ws (0.1)(2 contractions)(Dd) (ft) Free fall flow, Qfree 3.34 Ws' Du (3/2) (ft3/s) Submerged flow, Qsub Qfree (1 (Dd/ d/Du)3/2 )0.385 (ft3/s) Difference, (Qsub qsg), ft3/s Head difference between tank and channel, Hl 1 Du Dd (ft) Top of sluice gate set elevation, Els HGL2 Dd (ft) HGL at Point 1 (upstream of sluice gate), HGL1 HGL2 Hll (ft) Velocity head, HV1=0 (assume V 0) (m) EGL at point 1, EGL1 HV1 (m)

Min Day Avg Day


3.3 3.3 10.8 4

3.3 3.3 10.8 4

3.3 3.3 10.8 4

3.3 3.3 10.8 4

3.3 3.3 10.8 4

3.88

7.06

8.83

11.48

15.30

0.67

1.02

1.19

1.43

1.76

0.51 3.2

0.77 3.1

0.90 3.1

1.09 3.0

1.33 3.0

5.89

10.70

13.38

17.39

23.18

3.89 0.00

7.07 0.00

8.83 0.00

11.48 0.00

15.30 0.00

0.163

0.246

0.287

0.347

0.426

338.79 339.46

338.56

338.45

338.27

338.05

339.46 0.00 339.46

339.58 0.00 339.58

339.63 0.00 339.63

339.71 0.00 339.71

339.81 0.00 339.81

Maximum HGL1 (ft) Max HGL1 should equal HGL 6 for aeration tank.

TABLE 22.12

Peak

339.81

Example Hydraulic Calculation of a Typical Aereation Tank System Initial Operation Parameter

1. Wastewater flow rate, Q (ft3/s) (mgd) RAS flow, % of average flow (added downstream of aeration tank influent sluice gates) RAS flow, Qras Q RAS flow/100 (ft3/s) 2. Aeration tanks Total of nunber of units Number of units in operation Number of units on standby

Min Day Avg Day


Peak

35.31 23

56.50 37

70.63 46

113.01 73

113.01 73

20

50

50

100

100

11.30

28.25

35.31

70.63

70.63

3 2 1

3 2 1

3 3 0

3 3 0

3 2 1



22.112

Chapter Twenty-Two

TABLE 22.12


Parameter Flow rate per aeration tank in operation, q (ft3/s) Flow rate per aeration tank in operation including RAS flow (downstream of influent sluice gate), qras, (ft3/s)

Design Operation

Min Day Avg Day Avg Day Max Hour

Peak

17.66

28.25

23.54

37.67

56.50

23.31

42.38

35.31

61.21

91.82

339.79

339.79

339.79

337.79

339.79

340.12 330.28 23.31 19.69

340.12 330.28 42.38 19.69

340.12 330.28 35.31 19.69

340.12 330.28 61.21 19.69

340.12 330.28 91.82 19.69

0.50

0.75

0.66

0.95

1.25

340.63

340.87

340.79

341.08

341.37

0.09

0.13

0.11

0.17

0.22

340.71

341.00

340.90

341.24

341.59

23.31 19.7 196.9

42.38 19.7 196.9

35.31 19.7 196.9

61.21 19.7 196.9

91.82 19.7 196.9

320.94

320.94

320.94

320.94

320.94

19.69 5 984.3 0.06

19.94 5 984.3 0.11

19.85 5 984.3 0.09

20.14 5 984.3 0.15

20.44 5 984.3 0.23

0.35

0.52

0.46

0.67

0.88

0.013

0.013

0.013

0.013

0.013

6.56

6.59

6.58

6.61

6.64

Control point is located at Point 5 (aeration tank effluent weir). 3. At Point 6 Set maximum HGL6 effluent weir elevation0.33 (ft) Hydraulic calculations upstream of control point 4. Point 6 to Point 5 Headloss over sharp-crested weir Sharp-crested weir EL (ft) Effluent channel bottom EL (ft) Flow rate over weir, qras (ft3/s) Length of weir (ft) headloss, Hle5 (q/3.33L)(2/3) (ft) HGL at Point 5, HGL5 weir EL Hle5 (ft) Velocity head, HV5 V (gras/Wp/Hle / 5)2/2g (ft) EGL at Point 5, EGL5 HGL5 HV5 V (ft) 5. Point 5 to Point 4 Flow rate per aeration tank in operation, qras (ft3/s) Pass width, Wp (ft) Tank length, Lt (ft) Tank bottom elevation, ELtb avg. day WSEL 19.69 (ft) Water depth in tank at design average flow, Dt (ft) Number of passes per tank, Np Effective length of tank, L Lt Np (ft) Velocity, V V4 (ft/s) Critical depth, yc ((q2/g / /Wp2 ))(0.333) (ft) Friction headloss thruogh aeration tank channel Manning's n for concrete channel Hydraulic radius, R (Dt Wp)/(2 Dt Wp) (ft) Headloss, Hlf4 f (V4 V n/





Parameter 1.486 R (2/3))2 L (ft) Fitting headloss through a 90° bend Fitting headloss coefficient Kbend 1.0 Number of bends, Nb Headloss, Hlb4 Kbend V 2/2g (ft) V4 Velocity head, Hvsd (see below at Point 3) RAS flow, % of average flow EGL at Point 4, EGL4 EGL5 Hlf4 f Hvsd (ft) Velocity head, Hvsd (see below at point 3) HGL at point 4, HGL4 EGL4 HV4 (ft)

6. Point 4 to Point 3 Headloss over aeration tank influent sluice gates Sluice gate width, Ws (ft) Sluice gate heigth (ft) Flow per sluice gate, q (ft3/s) Upstream head over weir, Du (select so Zsub q 0) (ft) Downstream head over weir, Dd (q/3.33/Ws')(2/3) (ft) Effective sluice gate width, Ws' Ws (0.33)(2 contractions)(Du) (ft) Free fall flow, Qfree 3.34 Ws'Du (3/2) (ft3/s) Submerged flow, Qsub Qfree (1-(Dd/ d/Du)0.385 (ft3/s) Difference, (Qsub q), ft3/s (should de zero) Head difference between tank and channel, Hl4 Du Dd (ft) Velocity head downstream of sluice gate, HVsd (q/ Ws'/Dd / )2/2g, Velocity head upstream of sluice gate, HVsu (q/ Ws'/Du / )2/2g (ft) Top of sluice gate elevation, Els HGL4 Dd (ft) HGL upstream of sluice gate, HGLsu HGL4 Hl4 (ft) EGL upstream of sluice gate, EGLsu HGLsu HVsu (ft) Friction headloss through influent channel to tank #3 Average length of influent channel per tank, L3 Np Wp 3 tanks1/2 (ft) Influent channel width, W W3 (ft)

Design Operation


Peak

0.0000

0.0001

0.0000

0.0001

0.0003

1.0 8

1.0 8

1.0 8

1.0 8

1.0 8

0.0004

0.0014

0.0010

0.0030

0.0065

0.74

1.03

0.90

1.28

1.76

341.45 0.74 340.71

342.03 1.03 341.00

341.81 0.90 340.90

342.53 1.28 341.25

343.35 1.76 341.60

3.9 3.3 17.66

3.9 3.3 28.25

3.9 3.3 23.54

3.9 3.3 37.67

3.9 3.3 56.50

1.71

2.39

2.10

2.97

4.07

1.29

1.82

1.59

2.25

3.08

3.60

3.46

3.52

3.34

3.12

26.75

42.80

35.67

57.07

85.61

17.66 0.00

28.25 0.00

23.54 0.00

37.67 0.00

56.51 0.00

0.41

0.58

0.51

0.72

0.98

0.22

0.31

0.28

0.39

0.53

0.13

0.18

0.16

0.22

0.31

339.42

339.19

339.31

339.00

338.51

341.13

341.58

341.41

341.96

342.58

341.25

341.76

341.57

342.19

342.89

103.3

103.3

103.3

103.3

103.3

13.1

13.1

13.1

13.1

13.1



22.114

Chapter Twenty-Two

TABLE 22.12


Parameter Manning's n for concrete channel Influent channel bottom elevation, Elb ave.EGLsu 9.84 (ft) Water depth in influent channel, h3 HGLs Elb (ft) Hydraulic radius, R (h3 w3)/(2 h3 w3) (ft) Velocity, V3 V q/w3/h3 (ft/s) Headloss, Hlf3 f (V3 V n/1.486 R (2/3) )2 L3 (ft) Friction headloss through influent channel to tank #2 Flow rate, q2 2 q (ft3/s) Velocity, V2 V q/w2/h2 (ft/s) Headloss, Hlf2 (V2 n/1.486 R(2/3) )2 L3, m Friction headloss through influent channel to tank #1 Flow rate, q1 3 q (ft3/s) Velocity, V1 q/w1/h1 (ft/s) Headloss, Hlf1 f (V1 n/1.486 R(2/3))2 L3 (ft) HGL at Point 3, HGL3 HGLs Hlf3 f Hlf2 f Hlf1 f (ft) 7. Point 3 to Point 2 Headloss through sluice gate Sluice gate headloss coefficient Kgate 1.0 RAS flow, % of average flow (added downstream of aeration tank influent sluice gates) RAS flow, Qras Q RAS flow/100, cfs Sluice gate width, W W2 (ft) Sluice gate heigth, Hg (ft) Channel water depth, Dc (ft) Gate opening depth, Hg or Dc whichever is smaller (ft) Velocity through sluice gate, V5 V Q/W2 W (ft/s) Headloss, Hls2 Kgate V5 V 2/2g (ft) HGL at point 2, HGL2 HGL3 Hls2 (ft) 8. Point 2 to Point1 Allowance Exit headloss from primary sed, tank effluent pipe to aeration tank influent channel Primary effluent pipe diameter, Dp (ft) All PST effluent flow, Q 9 (ft3/s)

Design Operation


Peak

0.013

0.013

0.013

0.013

0.013

331.7

331.7

331.7

331.7

331.7

9.40

9.86

9.68

10.24

10.86

3.86 0.14

3.94 0.22

3.91 0.19

4.00 0.28

4.09 0.40

0.0000

0.0001

0.0000

0.0001

0.0002

35.31 0.29

56.50 0.44

47.09 0.37

75.34 0.56

113.01 0.79

0.0001

0.0002

0.00020.

0004

0.0008

35.31 0.29

56.50 0.44

70.63 0.56

113.01 0.84

113.01 0.79

0.0001

0.0002

0.0004

0.0009

0.0008

341.13

341.58

341.41

341.97

342.58

1.0

1.0

1.0

1.0

1.0

20

50

50

100

100

11.30

28.25

35.31

70.63

70.63

5.91 5.91 9.40

5.91 5.91 9.86

5.91 5.91 9.69

5.91 5.91 10.24

5.91 5.91 10.86

5.91

5.91

5.91

5.91

5.91

1.01

1.62

2.03

3.24

3.24

0.0159

0.0408

0.0637

0.1630

0.1630

341.14

341.62

341.47

342.13

342.75

6.56 35.31

6.56 56.50

6.56 70.63

6.56 113.01

6.56 113.01





Parameter Velocity, V1 (ft/s) Exit headloss coefficient Kexit 1.0 Exit headloss, hle1 (V12)/2/g / Kexit (ft) Friction headloss through PST effluent pipe section 2 Flow per pipe, Q (ft3/s) Pipe diameter, Dp2 (ft) Velocity, V12 (ft/s) Hazen–Williams coefficient, Cp Hydraulic radius, Rp2 Dp2/4 (ft) Length of pipe, Lp2 (ft) Slope, Sp2 [v12/(1.318 Cp Rp2(0.63))](1/0.54) (%) Headloss, hlf2 f Lp2 Sp2 (ft) Friction headloss through PST effluent pipe section 1 Flow per pipe, q (ft3/s) Pipr diameter, Dp1 (ft) Velocity, V11 (ft/s) Hazen-Williams coefficient, Cp Hydraulic radius, Rp1 Dp1/4 (ft) Length of pipe, Lp1 (ft) Slope, Sp1 [v11/(1.318 Cp Rp1(0.63))](1/0.54) (%) Headloss, hlf1 f Lp1 Sp1 (ft) Pipe entrance head loss Ke Head loss, hen1 Ke V112/2g (ft) HGL at upstream of PST effluent pipe, HGL1 HGL2 hle1 hlf2 f hlf1 f hen1 (ft) HGL7 of PST must be maximum of HGL1 (ft)

TABLE 22.13

Design Operation


Peak

1.04 1.0

1.67 1.0

2.09 1.0

3.34 1.0

3.34 1.0

0.0169

0.0434

0.0677

0.1734

0.1734

35.31 6.56 1.04 120.00 1.64 164.04

56.50 6.56 1.67 120.00 1.64 164.04

70.63 6.56 2.09 120.00 1.64 164.04

113.01 6.56 3.34 120.00 1.64 164.04

113.01 6.56 3.34 120.00 1.64 164.04

0.0001 0.0084

0.0001 0.0202

0.0002 0.0305

0.0004 0.0728

0.0004 0.0728

17.66 4.92 0.93 120.00 1.23 164.04

28.25 4.92 1.49 120.00 1.23 164.04

23.66 4.92 1.24 120.00 1.23 164.04

37.79 4.92 1.99 120.00 1.23 164.04

56.50 4.92 2.97 120.00 1.23 164.04

0.0001 0.0095

0.0001 0.0227

0.0001 0.0163

0.0002 0.0389

0.0005 0.0819

0.50 0.0067

0.50 0.0171

0.50 0.0120

0.50 0.0306

0.50 0.0685

341.18

341.73

341.60

342.44

343.14

343.14

343.14

343.14

343.14

343.14

Example Hydraulic Calculations of a Typical Multimedia Filter System Initial Operation Parameter

1. Wastewater flow rate, Q (ft3/s) (mgd) 2. Multimedia filters Total number of units Number of units in operation Number of units on standby Flow rate per operating multimedia filter, q (ft3/s)

Min Day Avg Day


Peak

35.3 23

56.5 37

70.6 46

113.0 73

113.0 73

6 4 2

6 5 1

6 5 1

6 6 0

6 5 1

8.83

11.30

14.13

18.83

22.60



22.116

Chapter Twenty-Two

TABLE 22.13


Parameter

Min Day Avg Day

Design Operation

Avg Day Max Hour

Peak

Hydraulic calculations at filter effluent 3. At Point 7 Max HGL in filtered water storage tank, HGL7 (ft) Velocity in storage tank, V V7 (ft/s) Max EGL in storage tank, EGL7 HGL7 V7 V 2/2g (ft)

323.72 0.00

323.72 0.00

323.72 0.00

323.72 0.00

323.72 0.00

323.72

323.72

323.72

323.72

323.72

4. At Point 6 Filtered water effluent channel weir Sharp-crested weir EL, Wel6 HGL7 0.33 (ft) Flow rate over weir Q (ft3/s) Length of weir (ft) Headloss, Hlw6 (q/3.33L)(2/3) (ft)

324.05 35.31 22.97 0.60

324.05 56.50 22.97 0.82

324.05 70.63 22.97 0.95

324.05 113.01 22.97 1.30

324.05 113.01 22.97 1.30

324.65

324.87

325.00

325.35

325.35

0.00

0.00

0.00

0.00

0.00

324.65

324.87

325.00

325.35

325.35

35.31 9.84 6.56 32.81 0.55

56.50 9.84 6.56 32.81 0.87

70.63 9.84 6.56 32.81 1.09

113.01 9.84 6.56 32.81 1.75

113.01 9.84 6.56 32.81 1.75

1.97 0.013

1.97 0.013

1.97 0.013

1.97 0.013

1.97 0.013

0.0003

0.0008

0.0012

0.0031

0.0031

3.3 8.83 1.04

3.3 11.30 1.34

3.3 14.13 1.67

3.3 18.83 2.23

3.3 22.60 2.67

0.0170

0.0278

0.0434

0.0772

0.1111

324.66

324.89

325.04

325.43

325.46

0.02

0.03

0.04

0.08

0.11

324.65

324.87

325.00

325.35

325.35

2.95

2.95

2.95

2.95

2.95

HGL at Point 6, HGL6 Wel6 Hlw6 (ft) Velocity in weir box, V V6, m (assume V 0) (ft) EGL at Point 6, EGL6 HGL6 V6 2/2g (ft) 5. Point 6 to Point 5 Loss through effluent concrete condiut Flow rate, Q (ft3/s) Width of condiut, Wc (ft) Depth of condiut, Dc (ft) Length of condiut, Lc (ft) Velocity, Vc (ft/s) Hydraulic radius, R Wc Dc/2/(Wc Dc) (ft) Manning's n Headloss, Hlc5 (Vc n/1.486 R(2/3) )2 Lc (ft) Exit loss from pipe to concrete conduit Effluent pipe diameter, Dp (ft) Pipe flow (for each filter) (ft3/s) Velocity, Vp (ft/s) Hle5 Vp 2/2g for sharp concrete outlet (ft) EGL at Point 5, EGL5 EGL6 Hlc5 Hle6 (ft) Velocity head at Point 5, HV5 V Vp2/2g (ft) HGL at Point 5, HGL5 EGL5 HV5 V (ft) 6. Point 5 to Point 4 Filter effluent pipe loss Pipe diameter, Dp (ft)





Parameter Max flow through filter effluent pipe q (ft3/s) Velocity of flow through pipe, Vp (ft/s) Hazen-Williams coefficient, Cp Hydraulic radius, Rp Dp/4 (ft) Length of pipe, Lp (ft) Slope, Sp [Vp/(1.318 Cp Rp0.63)](1/0.54) (%) Head loss, Hlf4 f Lp Sp (ft) Headloss through butterfly valve Kvalve (fully open) Valve diameter (ft) Headloss, hval4 Kvalve (Vp2/2g) (ft) Flow rate controller Venturi throat-to-inlet ration for long tube, Krate Inlet velocity, Vi V Vp (ft/s) Headloss, hrate Krate (Vi 2/2g) (ft) (minimum headloss when control valve is fully open) Pipe entrance loss Kent Headloss, Hlent Kent (Vp2/2g) (ft) EGL at Point 4, EGL4 EGL5 Hlf4 f Hval4 Hrate Hlent (ft) Velocity head, HV4 V V 2/2g, (assume V 0) (ft) V4 HGL at Point 4, HGL4 EGL4 HV4 V (ft) 7. Point 4 to Point3 Dirty filter head requirement, Hldf, f (ft) (assumed) (consult with filter manufacturer) Dirty filter HGL, HGLdf HGL4 Hldf (ft) Velocity head, HV3 V 0 (assume V3 V 0) (ft) Dirty filter HGL, HGLdf EGLdf HV3 V (ft) Clean filter headloss Filter bed area (ft2) Flow per filter, q (ft3/s) Filter rate, qfilt, (ft3/ min/ft2) Media depth, Dm (ft) Effective media size, Md (in) Headloss through filter, Hlf 2.32 ft loss per (ft3/ min/ft2)(consultant with manufacturer)

Min Day Avg Day

Design Operation

Avg Day Max Hour

Peak

8.83

11.30

14.13

18.83

22.60

1.29 120 0.74 49.21

1.65 120 0.74 49.21

2.06 120 0.74 49.21

2.75 120 0.74 49.21

3.30 120 0.74 49.21

0.0193 0.0095

0.0305 0.0150

0.0461 0.0227

0.0786 0.0387

0.1102 0.0542

0.30 2.95 0.0078

0.30 2.95 0.0127

0.30 2.95 0.0198

0.30 2.95 0.0353

0.30 2.95 0.0508

1.20 1.04 0.0203

1.20 1.34 0.0333

1.20 1.67 0.0521

1.20 2.23 0.0926

1.20 2.67 0.1333

0.50 0.0085

0.50 0.0139

0.50 0.0217

0.50 0.0386

0.50 0.0555

324.71

324.97

325.16

325.63

325.75

0.00

0.00

0.00

0.00

0.00

324.71

324.97

325.16

325.63

325.75

8.2

8.2

8.2

8.2

8.2

332.91

33.17

333.36

333.83

333.96

0.00

0.00

0.00

0.00

0.00

332.91

33.17

333.36

333.83

333.96

1722 8.83 0.308 3.28 0.2

1722 11.30 0.394 3.28 0.2

1722 14.13 0.492 3.28 0.2

1722 18.83 0.656 3.28 0.2

1722 22.60 0.787 3.28 0.2

0.7136

0.9134

1.1417

1.5223

1.8268



22.118

Chapter Twenty-Two

TABLE 22.13


Parameter

Min Day Avg Day

Entrance headloss through underdrain flume, Hlu 0.45 ft lossper (ft3/ min/ft2) (ft)(consult with filter manufacturer) Clean filter EGL, EGLcf EGL4 Hlf+ f Hlu (ft) Velocity head, HV3 V 0 (m) (assume V3 V 0) (ft) Clean filter HGL, HGLcf EGLcf HV3 V (ft) EGL requierd at Point 3, EGL3 EGLdf (ft) HGL requierd at Point 3, HGL3 HGLdf (ft) (Head requierd for dirty filter controls) 8. Point 3 to Point 2 Filter inlet discharge loss Keff Flow rate, q (ft3/s) Pipe diameter, Dp2 (ft) Velocity, Vp2 (ft/s) 2 headloss, Hld2 Keff (Vp2 /2g) (ft) EGLat Point 2, EGL2 EGL3 Hld2 (ft) Velocity head, HV2 V Vp22/g / (ft) HGL at Point 2 HGL2 EGL2 HV2 V (ft) 9. Point 2 to Point 1 Head loss through butterfly valve Kval (fully open) Headloss, Hlv1 Kval (Vp2 2/2g) Headloss through inlet pipe Length of pipe, Lp1 (ft) Hazen-Williams coefficient (Cp) Hydraulic radius, Rp Dp2/4 (ft) Headloss, Hlf1 f (Vp2/(1.318 Cp Rp1.63))(1/0.54) Lp (ft) Headloss through entrance to pipe Kent Headloss, Hlent Kent Vp 2/2g (ft) EGL at Point 1, EGL1 EGL2 Hlv1 Hlf Hlent (ft) Velocity head, HV1 0 (assume V1 0) (ft) HGL at point 1, HGL EGL1 HV1 (ft) Minimum required control HGL at Point 1 (ft) (max. HGL1 must equal HGL7 of final sedimentation tank)

Design Operation

Avg Day Max Hour

Peak

0.1384

0.1772

0.2215

0.2953

0.3543

325.56

326.06

326.52

327.45

327.94

0.00

0.00

0.00

0.00

0.00

325.56

326.06

326.52

327.45

327.94

332.91

33.17

333.36

333.83

333.96

332.91

33.17

333.36

333.83

333.96

1.0 8.83 3.0 1.29 0.0258

1.0 11.30 3.0 1.65 0.0423

1.0 14.13 3.0 2.06 0.0661

1.0 18.83 3.0 2.75 0.1176

1.0 22.60 3.0 3.30 0.1693

332.94 0.03

333.21 0.04

333.43 0.07

333.95 0.12

333.13 0.17

332.91

333.17

333.36

333.83

333.96

0.3 0.0078

0.3 0.0127

0.3 0.0198

0.3 0.0353

0.3 1.0508

65.6 120 0.74

65.6 120 0.74

65.6 120 0.74

65.6 120 0.74

65.6 120 0.74

0.0127

0.0200

0.0303

0.0516

0.0723

0.50 0.0129

0.50 0.0212

0.50 0.0331

0.50 0.0588

0.50 0.0847

332.97

333.27

333.51

334.10

334.33

0.00 332.97

0.00 333.27

0.00 333.51

0.00 334.10

0.00 334.33

334.33

334.33

334.33

334.33

334.33




Example Hydraulic Calculation of a Typical Cascade Aeration System Initial Operation Parameter

1. Wastewater flow rate, Q (ft3/s) (mgd) 2. Cascade aerator Total number of units Flow rate through aerator, Q (ft3/s) Optimal flow rate per ft width over step, q (ft2/s) DO concentration of postaeration influent, Co (mg/L) Desired DO concentration of postaeration effluent, Cu (mg/l)

Min Day Avg Day


Peak

35.3 23

56.5 37

70.6 46

113.0 73

113.0 73

1 35.31

1 56.50

1 70.63

1 113.01

1 113.01

0.7029

0.7029

0.7029

0.7029

0.7029

0.00 5.00

0.00 5.00

0.00 5.00

0.00 5.00

0.00 5.00

16.4 2.15

16.4 3.44

16.4 4.31

16.4 6.89

16.4 6.89

0.524 3.9

0.717 3.9

0.832 3.9

1.138 3.9

1.138 3.9

16.93

23.16

26.87

36.74

36.74

13.55

18.53

21.49

29.39

29.39

0.93

0.93

0.93

0.93

0.93

0.42 1.53

0.36 1.43

0.33 1.39

0.28 1.32

0.28 1.32

Calculation of aerator dimensions with with predetermined weir length 3. Weir length, W (ft) Flow over weir, q Q/W (ft3/s/ft) Critical depth at upstream step edge, hc (q2/g)1/3 (ft) Optimal fall height of nappe, h (ft) Length of downstream bubble cushion, Lo 0.0629(h0.134)(q0.666) (ft) Length of downstream receiving channel, L 0.8Lo (ft) Optimal tailwater depth, H' 0.236 h, ft for h 3.9 ft Deficit ratio log at 68 F, Inr68 1.86(h1.31)(q0.363)(H H0.31) Deficit ratio, r20 Calculate concentration of dissolved oxygen downstream of step. If concentration is less than desired concentration, add another step and again calculate DO downstream concentration. Continue adding steps until the desired DO oncentration is achieved. Select cascade aerator dimension corresponding to those calculated for average flow. 4. Calculation of number steps to obtain desired DO Desired DO concentration at average flow, Cu (mg/L) Step 1 effluent DO, C1 9.07(1 (1/r20)) Co/r20) (mg/L) Step 2 effluent DO, C2 9.07 (1 (1/r20)) Co/r20) (mg/L)

5.00 3.13

2.73

2.55

2.20

2.20

4.80

4.51

4.38

4.13

4.13



22.120

Chapter Twenty-Two

TABLE 22.14


Step 3 effluent DO, C3 9.07 (1 (1/r20)) Co/r20) (mg/L)

Min Day Avg Day


Peak

6.00

5.79

5.70

5.52

5.52

3.28 3.94

3.28 3.94

3.28 3.94

3.28 3.94

3.28 3.94

HGL at Point 1, HGL1 (ft)

319.98

319.98

319.98

319.98

319.98

HGL at Point 2, HGL2 HGL1 h (ft)

316.04

316.04

316.04

316.04

316.04


312.11

312.11

312.11

312.11

312.11


308.17

308.17

308.17

308.17

308.17

In this example, the desired downstream DO concentration for average flow is achieved after three steps. 5. Calculation of HGL at each step Head loss from filtered water storage tank to point 1 (ft) Cascade fall height, h (ft)


Chapter 22 - Water and Waste Water Treatment Plant Hydraulics

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