Hydraulics of Sewage Treatment Plants Sec-6

Section 6

Sect Sectio ion n 6: 6: 6.1

Hydraulics of Sewage Treatment Plants

6 -1

Hydr Hydrau auli lics cs of Sewa Sewage ge Tr Trea eatm tmen entt Pla Plant ntss

Introduction

In the design of many sewage treatment plants, the hydraulics have frequently been given scant regard. This often leads to severe operational difficulties such as component units overflowing under peak conditions and component units not coping if some are out of service for maintenance purposes. The efficient operation of sewage treatment plants requires an understanding of the hydraulic aspects of the various flow processes occurring. Treatment plants utilise unit operations and unit processes to achieve the desired degree of purification. In unit operations, the treatment or removal of contaminants is brought about by physical forces. On the other hand, in unit processes, the treatment occurs through chemical and biological reactions. This chapter does not provide full details on the design and operation of a sewage treatment plant. Such details may be found in other texts. This chapter is concerned with the hydraulic design of sewage treatment plants. Although this is often seen as a challenging exercise, the hydraulic principles involved in individual units are normally reasonably basic. The challenge lies in understanding how individual units interact hydraulically with each other. Broadly, the aims of this chapter are two-fold: 1) To identify the hydraulic principles associated with various unit operations and processes. 2) To understand how knowledge of the hydraulics leads to improved system design. The following section presents a discussion of the hydraulic aspects in broad outline. In later sections, the hydraulics of individual unit operations and processes are studied and the interaction among the various units is studied. Finally, the concept of the complete hydraulic profile is considered in some detail. 6.2

Broad Concepts

For each unit process and unit operation, the hydraulic calculations will require the application of one or more of the fundamental concepts, developed in Chapter 1 of this Manual. A typical example is shown in Figure 6.1 which shows the unit process hydraulics for a clarifier. Calculation of the water surface elevation difference between the effluent manhole and the clarifier would require the use of the following principles:

•

Pressure flow equation and pipe fitting equation for determining the head loss in the pipe between the manhole and the clarifier.

©2000 Assoc.Prof R.J.Keller

Manual of Practice Hydraulics of Sewerage Systems

Section 6


6 -2

•

Side overflow weir equation for calculating the highest water surface elevation in the effluent launder of the clarifier.

•

V-notch weir equation for calculating the head on the weir crest.

Moving further upstream from the clarifier, the engineer may need to consider the head loss across the influent ports into the clarifier, and the head loss in the influent pipe.

Figu Figure re 6.1: 6.1:

Schem Schemati aticc of Hydr Hydrau aulic licss for a Typ Typica icall Clari Clarifie fierr

For each unit process, the designer must understand how the sewage flows through and what water depths are required for the process. Within each process, various means are used to distribute flow, maintain a certain water depth, and control the flow. Such means include weir gates, valves, weirs, baffles, orifices, orific es, launders and under-drains. Each of these imposes a head loss on the system and must be considered in the h ydraulic calculations. Each unit process, its respective flow devices, and interconnecting piping must be carefully analysed. As a consequence, the water surface elevations can be calculated and the structure elevations and pumping needs can be established. This information can be summarised and presented in the form of the hydraulic profile through the entire sewage treatment plant. In carrying out the unit process hydraulic calculations, the designer should consider the need to control and equally distribute the flow into multiple tanks or within a single tank. Wherever possible, static devices - such as distribution boxes, channels, weirs, and header pipes - are better suited than dynamic devices. The latter include modulating ports, gates, and valves. Each requires a control system which has inherent disadvantages of potential failure and high maintenance. Each unit process has particular hydraulic characteristics that should be addressed. In the following sections, some of the major issues for ©2000 Assoc.Prof R.J.Keller


Section 6


6 -3

consideration in the unit processes of sewage treatment plants are discussed. The final section deals with the development of the hydraulic profile. 6.3

Principles of Sedimentation 6.3.1 3.1

Prel relimi iminary

Sedimentation is the separation of suspended particles, heavier than the wastewater, by gravity. It is one of the most widely used unit operations in sewage treatment plants. In primary treatment, sedimentation is the main unit process and is used for grit removal and removal of other particulate matter. It is responsible for removing 50-70% of suspended solids. The removed suspended solids contain between 25 and 40% of the the BOD. Following biological (secondary) treatment, sedimentation is used to remove the biological floc in the activated sludge settling basin and for solids concentration in sludge thickeners. In most cases, the purpose of secondary sedimentation is to produce a clarified effluent which may be directly discjarged into inland waterways. Where it is used for solids concentration, the aim is to produce a sludge which can be easily handled and treated. An understanding of the principles of sedimentation is necessary for the effective design of sedimentation tanks. Within such tanks, three processes may take place as follows: Sedimentation, defined as the removal of particle s by settling under gravity. Clarification, which is similar to sedimentation but refers specifically to the removal of suspended matter to give a clarified effluent. Thickening, in which settled impurities are concentrated and compacted on the floor of the tank and in the sludge collecting hoppers. In this section, the different classes of sedimentation are identified. The hydraulics involved in each is then discussed and outline analyses presented. In later sections of this chapter, the use of these concepts in design is discussed. 6.3. 6.3.2 2

Clas Classe sess of Sed Sedim imen enta tati tion on

Sedimentation is classified classifie d according to the nature nat ure of the particles to t o be removed and their concentration. Individual particles may be discrete, such as sand and grit; or flocculent, such as organic materials and biological solids. Particle concentrations may vary from very low, through moderate, to high concentrations in which adjacent particles are in contact. Commonly, four classes are identified and these are summarised in Table 6.1.



Section 6


Sedimentation Class

Description

Class 1

Sedimentation of particles in suspension of low solids concentration. Particles settle individually without interaction with neighboring particles.

(Discrete particle settling)

Class 2 (Flocculent settling)

6 -4

Application

Removal of grit and sand particles from sewage.

Dilute suspension of Removes some particles which flocculate suspended solids in during the sedimentation primary settling units operation. Flocculation and in upper parts of causes the particles to secondary units. increase in mass and Removes chemical settle at a faster rate. flocculent in settling tanks.

Class 3

In suspensions of intermediate (Hindered settling and concentration, interzone settling) particle forces hinder the settling of neighboring particles. The mass of particles tends to settle as a unit with individual particles remaining in fixed positions with respect to each other. A solids-liquid interface develops at the top of the settling mass.

Process often occurs in secondary settling units used in conjunction with biological treatment facilities

Class 4

The concentration of Usually occurs within particles is so high that a the lower layers of a structure is formed. sludge mass. It occurs Further sedimentation can at the bottom of deep only occur through secondary settling units. compaction of the It is particularly structure. Compaction important in activatedtakes place through the sludge final settling weight of the particles tanks where the which is continuously activated sludge must increased by be thickened for sedimentation from the recycling to the aeration over-lying liquid. tanks.

(Compression Settling)

Table 6.1:


Classes of Sedimentation Phenomena


Section 6

6.3.3


6 -5

Class 1 Sedimentation

Because the particles are considered to settle independently of neighboring particles, Class 1 sedimentation can be analysed with reference to a single particle. The terminal velocity of a discrete particle settling in a fluid is reached when the drag force, associated with the motion of the particle, is equal to the submerged weight of the particle. For a particle of diameter d , density ρ , falling at a terminal velocity v p, in a fluid of density ρ f , the submerged weight, W , is given by: W = ( ρ − ρ f ) g

π d 3 6

(6.1)

The drag force on the particle is given by: F D = C D

π d 2 1 4 2

ρ v p2

(6.2)

The equilibrium condition is reached when W is equal to F D – ie:

( ρ − ρ ) g f

πd 3 6

= C D

π d 2 1 4 2

ρ vp2

(6.3)

Re-arrangement of Equation (6.3) yields:

v p =

4 gd ( ρ − ρ f ) 3 C D

ρ

(6.4)

The drag coefficient, C D, is not constant but varies with Reynolds Number and particle shape. Furthermore, the particle diameter and density are usually not known and the particles are irregular in shape. This means that Equation (6.4) cannot normally be used in practice. Despite this, Equation (6.4) does show that the terminal velocity, v p, is dependent on particle and fluid properties and this is of value in understanding sedimentation behaviour. Furthermore, it is known that the termi nal velocity in practice is reached very quickly. Consequently, for non-flocculent particles and uniform fluid flow, the settling velocity is effectively constant throughout the settling time. In the following, this concept is applied to settling in an ideal sedimentation tank. It is shown that this leads to identification of an important design parameter, the surface loading rate. Three common types of sedimentation tank are shown schematically in Figure 6.2. These are classified as (a) Rectangular Horizontal Flow Tanks, (b) ©2000 Assoc.Prof R.J.Keller


Section 6


6 -6

Circular Radial Flow Tanks, and (c) Upflow Tanks. In each, four zones may be identified as follows: Inlet Zone:

In which momentum is dissipated.

Settling Zone: In which quiescent settling occurs as the water flows towards the outlet. Outlet Zone:

In which the flow converges upwards to the decanting weirs or launders.

Sludge Zone: In which settled material collects and is removed by sludge hoppers.

Figure 6.2:

Schematics of Different Types of Settling Tank

Considering first the rectangular horizontal flow tank of Figure 6.2, it is evident that the critical particle for design purposes is that which enters the tank at point A and settles at the end of the tank at point B. This particle moves through the tank with a horizontal velocity component of V h and a vertical component of V p (the terminal velocity). Noting that the effective length and height of the tank are respectively L and H , the time required for the particle to settle is given by: t =

Now, V h =

Q

BH (6.5) yields:

H V p

=

L V h

(6.5)

, where B is the width of the tank. Substitution into Equation

V p =

Q BL

It is clear that BL is equal to the tank surface area, A, so that:



(6.6)

Section 6


V p =

Q A

6 -7

(6.7)

Equation (6.7) states that the slowest moving particles which could be expected to be completely removed in an ideal sedimentation tank would have a settling velocity of Q/A. This parameter is called the surface loading rate or overflow rate and is a fundamental parameter governing sedimentation tank performance. A similar analysis may be developed for the circular radial flow tank as follows: With reference to Figure 6.2, the detention time is given by: t =

Now, V r =

Q

2π rH

H V p

=

R 2

dr

R1

V r

(6.8)

and substitution into Equation (6.8) yields:

t =

2π H Q

R2 R1

rdr

(6.9)

Evaluation of the integral leads to: t =

π ( R22 − R12 ) H Q

(6.10)

Now, π ( R22 − R12 ) is equal to the surface area, A. Substituting into Equation (6.10), and noting that t =

H V p

from Equation (6.8):

V p =

Q

(6.11)

A

which is identical to Equation (6.7). Considering now the upflow tank of Figure 6.2, it is clear that the minimum upflow velocity, V u, is equal to Q/A. The limiting case for particle removal occurs when V u=V p, from which: V p =

Q A

(6.12)

which is identical to Equations (6.7) and (6.12). Ideally then, all particles with a settling velocity greater than Q/A will be completely removed from the fluid. Additionally, however, it is evident that ©2000 Assoc.Prof R.J.Keller


Section 6


6 -8

for tanks type (a) and (b), particles with lower settling velocities of v p /n will be removed in the proportion 1/ n. It should be noted, however, that in an upflow tank of type (c), no particles with settling velocities less than Q/A can be removed. 6.3.4


Under quiescent conditions, suspended particles exhibit a natural tendency to flocculate. The settling characteristics of flocculating sediments are different from those of Class 1 sediments because the various sized particles settle at different rates. As larger, faster-settling particles overtake slower settling particles, they may collide and flocculate, forming larger aggregates with an increased settling velocity. Thus, the typical path followed by such flocculant particles is curved. The situation is shown schematically in Figure 6.3.

Figure 6.3:

Effect of Tank Depth on Removal of Class 1 and Class 2 Particles

It is an important requirement of sedimentation tanks for flocculent suspensions that sufficient depth is available to provide the opportunity for particle aggregation to occur. This becomes clear through an examination of Figure 6.3, which compares the behaviour of Class 1 and Class 2 sedimentation if the tank depth is reduced. For the tank shown with a depth of H , path ACB represents the settling path for a critical Class 1 sediment, and path ADB that for a flocculent particle. For the latter, the instantaneous settling velocity is the tangent to the curve. Now, consider the effect of reducing the depth of the tank to H/2. The forward velocity will be doubled and the total time of travel through the tank will be halved. The settling path followed by the critical Class 1 sediment will now be AX1, while that of the Class 2 sediment will be AY 1.



Section 6


6 -9

Thus, it can be seen that the critical Class 1 sediment will stil just reach the bottom of the settling zone. The Class 2 sediment, however, will not have reached the tank floor and will be drawn off in the tank effluent. It is evident that the minimum average settling velocity for particles to be removed is the surface loading rate. However, by comparison with Class 1 sedimentation, removal of Class 2 sediments depends on the depth or detention time provided, in addition to the surface loading rate. Now, the detention time, t , is given by: t =

Tank volume Q

(6.13)

Then, for a rectangular tank: t =

BLH Q

=

H Q A

(6.14)

Equation (6.14) demonstrates that if any two of the three parameters detention time, depth, and surface loading rate are given, the third is fixed. Ideally, the effects of depth and detention time on solids concentration is obtained by examining representative samples obtained at various depths. These are not usually available, especially for new schemes, and use is made of standard values. Class 2 sedimentation removes a portion of the suspended solids in untreated sewage in primary clarifiers, suspended solids in the upper portions of secondary clarifiers, and the chemical floc in settling tanks. 6.3.5


Class 3 sedimentation is associated with an increased concentration of particles in the suspension. A condition is eventually reached where the particles are so close together that the velocity fields of the fluid displaced by adjacent particles overlap. Additionally, there is a net upward flow of liquid displaced by the settling particles, resulting in a reduced particle settling velocity. For this reason, Class 3 sedimentation is frequently called “hindered” settling. Most commonly, hindered settling occurs in the extreme case where the very high particle concentration causes the whole suspension to settle as a blanket. Under these conditions, several distinct zones may be observed, separated by concentration discontinuities, and this leads to the descriptive term of “zone” settling. Figure 6.4 shows a typical batch settling column test on an activated sludge. The slope of the settling curve represents the settling velocity of the interface between the suspension and the clarified liquid. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -10

Class 3 sedimentation frequently occurs in secondary settling clarifiers used in conjunction with biological treatment facilities. In designing such clarifiers, the major design parameter is the surface loading parameter because, if the surface loading parameter is greater than the zone settling velocity, solids will be carried out by the effluent. Design applications for clarifiers where Class 3 sedimentation may occur are considered in a later section.

Figure 6.4: 6.3.6

Suspension Exhibiting Hindered Settling Behaviour


Class 4 sedimentation is characterised by particle concentrations which are so high that adjacent particles are actually in contact with each other. Consequently, a structure is formed and further set tling can only occur through compression of the structure. Compression takes place through a continuous increase in the weight of overlying particles. These are constantly added to the structure by sedimentation from the supernatant liquid. Under the increased load, the void spaces in the structure are gradually diminished and water is squeezed out of the matrix. Class 4 sedimentation usually occurs in the lower layers of a deep sludge mass, for example in the bottom of deep secondary settling facilities, and in sludge thickening facilities. It is particularly important in activated sludge final settling tanks where the activated sludge must be thickened for recycling to the aeration tanks. ©2000 Assoc.Prof R.J.Keller


Section 6

6.4


6 -11

Hydraulics of Screens 6.4.1

Preliminary

Screening of sewage is one of the oldest treatment processes. The purpose of screens is to remove gross pollutants from the sewage stream to protect downstream operations and equipment from damage. For this reason, it is normally the first unit operation used at sewage treatment plants. Screens are classified as primary screens, secondary screens, and microstrainers. In this section, each type of screen is defined and its role discussed. The hydraulic aspects are then presented. Hydraulic design equations are then developed and their use in practice illustrated by examples. 6.4.2

Primary Screens

Primary screens are typically located at the inlet to sewage treatment plants and also at the inlet to pumping stations. They are designed to remove coarse debris such as rags, solids, and sticks which could cause damage by fouling pump impellers or interfering with downstream performance in sewage treatment plants. Primary screens are normally classified as coarse with openings of 50-150 mm or medium with openings 20-50 mm. Fine screens are typically secondary screens and are considered later. There are several factors that need to be taken into account in screen design. These include the strength of the screen material and its resistance to corrosion, the clear screen area, the maximum flow velocity through the screen to prevent dislodging of screenings, the minimum velocity in the approach channel to prevent sedimentation of suspended matter, and the head loss through the screen. The analysis of a primary screen involves the determination of the head loss across it. The head loss is primarily a function of the flow velocity and the screen openings, but may also be dependent on bar size, bar spacing, and the angle of the screen from the vertical. Several equations have been developed, but only those most widely used are considered herein.

Figure 6.5: ©2000 Assoc.Prof R.J.Keller

Schematic of Sloping Bar Screen


Section 6


6 -12

Figure 6.5 shows a schematic of a sloping bar screen. Application of Bernoulli’s equation yields: h1 +

where h1

v12

2 vsc

= h2 +

2g

+ losses

2g

(6.15)

is the upstream depth of flow

h2

is the downstream depth of flow

g

is the acceleration due to gravity

v1

is the upstream velocity

vsc

is the velocity through the screen

For a clean or partially blocked screen, the losses are usually incorporated into a coefficient and Equation (6.15) is expressed as: losses = ∆h = h1 − h2 =

where C d

1

(v 2gC 2 d

2 sc

− v12 )

(6.16)

is a discharge coefficient with a typical value of 0.84.

Alternatively, an orifice equation may be applied in t he form:

∆h = where Q A

vsc2

=

2 gCd2

1



Q

2 g Cd A

2



(6.17)

is the flow rate is the effective open area of the submerged screen

It should be noted that the discharge coefficient in Equation (6.17) is different from that in Equation (6.16). In the latter equation, the value of C d is dependent on screen design parameters and is supplied by the screen manufacturer or by experimentation. If the screens are to be manually cleaned, the effective open area should be taken as 50 % of the actual open area, representing the half-clogged condition. The head loss should be estimated under conditions of maximum flow. If the bar screen is clean, Kirschmer’s equation may be used for estimating the head loss as follows:

∆h = β  where β

W b



1.33

hv sinϑ

is a bar shape factor, as given in Table 6.2



(6.18)

Section 6


W

is the total transverse width of the screen

b

is the total transverse clear spacing between bars v12

hv

is the upstream velocity head  =

θ

is the angle of the bars to the horizontal

2g

6 -13



Bar Type

β

Sharp-edged rectangular

2.42

Rectangular with semicircular upstream face

1.83

Circular

1.79

Rectangular with downstream face

semicircular

upstream

Tear shape

and

1.67

0.76

Table 6.2:

Bar Shape Factor for Kirshmer’s Equation

It should be noted that Kirshmer’s equation is a general form of the standard head loss equation:

∆h = K where v K

v2

(6.19)

2g

is identified as v1 is given by K = β 

W b



1.33

sinθ

It should be noted that the expressions developed above are of use in determining the minimum energy losses through screens , but are of little value in determining the energy loss once material begins to accumulate behind the screen. Design should take into account the maximum increase in head loss likely to occur under the conditions of maximum flow rate and minimum cleaning frequency. It is especially important with manually raked screens that sufficient freeboard is provided in the upstream channel to avoid the danger of overtopping at high flows. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -14

Example 6.1

A mechanically cleaned wastewater bar screen is constructed using 6.5 mm wide bars with a clear spacing of 5.0 cm. The wastewater flow velocity in the channel immediately upstream of the screen will vary from 0.4 m/sec to 0.9 m/sec. Determine the design head loss for the screen at the two extremes of flow. (Assume that the discharge coefficient has a value of 0.84.) Solution:

Head Loss =

1

(v 2 gC 2 d

2 sc

− v12 )

If v1 is given, vsc can be calculated, knowing the screen geometry. Continuity: v1h1 w1 = vsc h1 wsc ( clear ) w1 wsc( clear )

=

=

bar spacing + bar width bar spacing

50 + 6.5 50

= 1.13

∴ vsc = 1.13v1 ∴ ∆h =

1 2 x9.81x( 0.84 ) = 0.02v1

2

x(113 . 2 v12 − v12 )

2

v1 = 0.4 m/sec

∆h = 3.2 mm

v1 = 0.9 m/sec

∆h = 16.2 mm

Primary screens may be manually cleaned or mechanically raked. Manually cleaned screens are only fitted in small treatment plants, typically servicing a population equivalent (PE) of less than 5,000. Mechanically raked screens are recommended for all plants servicing a PE greater than 2,000. Figure 6.6 shows a schematic of a manually raked screen. The maximum clear spacing between bars is typically set at 25 mm, although American practice ©2000 Assoc.Prof R.J.Keller


Section 6


6 -15

permits spacings up to 50 mm. To facilitate cleaning, the bars are normally set 0 at 30 – 45 from the vertical. The screenings are manually raked on to a perforated plate where they drain, prior to removal. Cleaning must be frequent to avoid clogging. Infrequent cleaning may result in significant upstream backwater caused by he buildup of solids. When cleaning is carried out, the sudden release of the ponded water leads to flow surges.

Figure 6.6:

Schematic of Manually Raked Screen

A schematic of a mechanically raked bar screen is shown in Figure 6.7. Typically, the maximum clear spacing between bars is 25 mm, although American practice permits spacings up to 38 mm. A spacing of 18 mm is considered satisfactory for the protection of downstream equipment.

Figure 6.7:


Schematic of Mechanically Raked Bar Screen


Section 6


6 -16

0

Mechanically raked screens are normally set at between 0 and 45 from the vertical. The use of such screens leads to reduced labour costs, improved flow conditions, and improved capture of screenings. A large number of proprietary screens with mechanical rakes are available. Manufacturers will normally provide design charts to facilitate selection of the correct screen size for a particular service. Figure 6.8 shows a schematic of another type of screen – a drum screen. Screenings naturally fall from the screen as it rotates above the hopper. A water spray assists in removing screenings.

Figure 6.8:

Schematic of Drum Screen

The velocity in the approach channel is normally kept between about 0.3 m/sec and 1 m/sec. The lower limit is designed to prevent the settling of coarse matter while the upper limit is designed to prevent the screens being carried away by the flow. An example illustrating the design technique for a screen and screen chamber is presented in Example 6.2. Example 6.2

Design a screen and screen chamber and determine its hydraulic characteristics for a loading of 10,000 PE. All material larger than 12 mm is to be screened out. The screen is a bar screen with rectangular bars of 5 mm transverse dimension. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -17

Note: At the peak design flow, the velocity through the screen should be 0.9 m/sec The water level downstream of the screen is controlled by a downstream long-throated flume which gives a depth of 400 mm at the peak design flow and 175 mm at ADWF. In particular, a.)

Determine head loss across screen

b.)

Determine screen chamber width

c.)

Check velocities

d.)

If the screen is 50 % blocked, calculate the head loss across it.

Solution:

Estimate loads ADWF = 225 l/day/PE Peak flow factor = 4.7 × (PE)

-0.11

(PE in thousands)

Load = 10,000 PE

∴SDWF

= 2.25Ml/day = 26l/sec -0.11

= 4.7 × 10

Peak flow factor

= 3.65

∴Peak flow

= 3.65 × 26 = 95l/sec

Bar spacing

= 12mm (will screen out all larger material)

Bar thickness = 5mm



Section 6


If screen velocity is 0.9m/sec for peak flow, calculate v1 v 1 = v sc ×

= 0.9 ×

bar spacing bar spacing × bar width

12 17

= 0.64m/sec

a.)

Determine head loss

h2 =

=

1 2 d

2 gC

( v sc2 − v12 )

1 2 × 9.81 × 0.84

2

( 0.9 2 − 0.64 2 )

= 0.029m ∴Depth upstream of screen = 400(mm) + 0.029(m) = 429mm b.)

Determine screen chamber width.

From continuity, required clear screen width ( W sc( clear ) ) is Q = h1 × Wsc ( clear ) × v sc

∴ W sc( clear ) =

0.095 0.429 × 0.9

= 0.246m

∴Required screen chamber width = 0.246 ×

17 12

= 0.349m or 350mm

(CHECK against approach velocity) ©2000 Assoc.Prof R.J.Keller


6 -18

Section 6


v1 =

Q W × h1

c.)

6 -19

0.095

=

0.349 × 0.429

=0.64m/sec

Check velocities 3

ADWF = 0.026m /sec Associated h2=175mm 0.026

∴ v2 =

× 0.349 0175 .

= 0.426m / sec Now, because the flow is lower, we would expect a reduced head loss as well.

∴The upstream depth will be less than 0.175 + 0.029 < 0.204m

∴ v1 >

0.026 0.204 × 0.349

= 0365 . m / sec

>0.3m/sec

∴ O.K. Note: We could calculate v1 exactly, but the above argument removes the need to do so. d.)

Head loss with screen half blocked

Energy equation: h1 +

v12

2g

= h2 +

v 22

2g

+ h L 3

For peak flow Q = 0.095m /sec h2 = 0.4 m

∴ h L =

v1 =

1 2 × 9.81 × ( 0.84) Q

h1 × 0.35


=

2

( vsc2 − v12 )

0.095 h1 × 0124 .

=

0.766 h1


Section 6


6 -20

Substitute for v1 , h2 , v2 , v sc in energy equation h1 +

( 0.271) 2 19.6 h12

∴ h1 +

= 0.4 +

0.00375

( 0.679) 2 19.6

+

1 19.6 × 0.84 2



0.766 2 h12

−

0.2712 h12



− 0.4235 = 0

2

h1

Solve by trial h1 = 0539 . m

∴Head loss

= 539 – 400

= 139mm ∴ v sc =

Q

=

h1 0124 .

0.095

× 0.539 0124 .

=1.42m/sec v1 =

6.4.3

0.271 h1

=

0.271 0539 .

= 0503 . m / sec

Secondary Screens

Secondary screens have smaller openings than primary screens and are installed following pumping and ahead of the grit chamber. Their purpose is to remove material such as paper, plastic, cloth, and other particles which may affect the treatment process downstream; and to minimise blockages in sludge handling and treatment facilities. Secondary screens are analysed in the same way as primary screens. The only difference is in the maximum clear spacing of bars. This is typically around 12 mm, although openings as small as 6 mm have been used in practice. 6.4.4

Microstrainers

Microstrainers have been used to further reduce suspended solids in effluent from secondary clarifiers following biological treatment. They typically comprise very fine fabric or screen wound around a drum. They are typically about 75 % submerged and rotate with wastewater flowing from inside to outside. Microstrainer openings are typically from 20 – 60 µm. They are successful at removing suspended solids, but not bacteria.



Section 6


6 -21

The main hydraulic aspect is the determination of the head loss, which is analysed semi-empirically. It is observed that the head loss is directly proportional to flow rate, degree of clogging, and time; and inversely proportional to the surface area of the strainer. These observations lead to: dh dt

where k

= k

Q A

(6.20)

h

is a characteristic loss coefficient.

Integration of Equation (6.20) leads to: h = h0 e

where h0

Q k t A

(6.21)

is the head loss across the clean strainer.

The United States Environmental Protection Agency surveyed a number of microstrainers treating secondary effluent with solids concentrations in the range of 6 – 65 mg/L and found average removals of between 43 and 85 %. Typical design parameters are presented in Table 6.3. Property

Typical Value

Screen Mesh

20 – 25 µm

Submergence

75 % of height

Hydraulic Loading

12 – 24 m /m /h

Head Loss

7.5 – 15 cm

Maximum Head Loss

30 – 45 cm

Peripheral Drum Speed

4.5 m/min at head loss of 7.5 cm

3

2

40 – 45 m/min at head loss of 15 cm Typical Drum Diameter

Table 6.3: 6.5

3m

Typical Microstrainer Design Parameters

Hydraulics of Grit Chambers 6.5.1

Preliminary

Within sewage treatment plants, grit - comprising sand, egg shells, coffee grounds and other non-putrescible material – may cause severe problems in



Section 6


6 -22

pumps, sludge digestion facilities, and de-watering facilities. In addition, it may settle out in downstream pipes and processes. The grit removal process is carried out at an early stage of treatment because the grit particles cannot be broken down by biological processes and the particles are abrasive and wear down the equipment. Because the grit material is non-putrescible, it requires no further treatment following removal from the sewage treatment process and ultimate disposal. It should be noted, however, that the location of grit chambers upstream of the sewage pumps at the entrance to the sewage treatment plant, would normally involve placing them at a considerable depth involving substantial expense. It is, therefor, usually more economical to pump the sewage, including the grit, to grit chambers located at a convenient position upstream of the treatment plant units. It is recognised that the pumps may require greater maintenance as a result. Grit chambers are designed to remove inorganic solids of size greater than about 2 mm. Removal is commonly effected using settlement, separation using a vortex, or settlement in the presence of aeration. (In the latter process, aeration keeps the lighter organic particles in suspension.) There are important hydraulic principles associated with each of these three processes. In this section, the choice of grit removal process is first discussed. The three main types of grit chamber are then described and the hydraulic aspects of the operation of each are described qualitatively and, where appropriate, quantitatively. Design aspects are also discussed. 6.5.2

Choice of Grit Removal Process

The choice of grit removal process depends largely on the size of the sewage treatment plant. For a PE less than 5,000, a horizontal flow (constant velocity) settling chamber is commonly used. For medium-sized treatment plants, handling a PE of between 5,000 and 10,000, a vortex type grit chamber is commonly used. For plants handling a PE greater than 10,000, the aerated grit chamber is often specified, although the vortex type chamber may also be used. Whichever type is used, it is vital that the unit must operate effectively over the full range of expected flows. Other non-hydraulic considerations include grit removal from the unit, which may be manual or mechanical; handling, storage, and disposal of grit; and the provision of standby or bypass facilities. 6.5.3

Horizontal Constant Velocity Grit Chamber

The horizontal flow grit chamber is basically an open channel with a detention time sufficient to allow design particles to settle. Additionally, the velocity



Section 6


6 -23

must be sufficiently high that organic materials are scoured so that they pass through the grit chamber for subsequent biological treatment. The Camp-Shields equation is commonly used to estimate the scour velocity required to re-suspend settled organic material. This equation is expressed as: vs =

8kgd ρ p − ρ f





ρ

(6.22)

where vs

is the velocity of scour

d

is the particle diameter

k

is an empirical constant (typically 0.04 – 0.06)

f

is the Darcy-Weisbach friction factor (typically 0.02)

ρ p

is the particle density

ρ

is the fluid density

Typically, this equation yields a required horizontal f low velocity of 0.15 – 0.3 m/sec. This compares well with the Malaysian design standard of 0.2 m/sec. The primary hydraulic design issue for the horizontal flow grit chamber is the maintenance of the constant velocity in the channel, despite large variations in the flow rate, based on a typical diurnal flow pattern. The problem is illustrated in the following. Consider a rectangular channel with the flow passing over a rectangular weir. The discharge relationship for the weir is: 3

Q = Cd B 2 gH 2

where C d

(6.23)

is a discharge coefficient

B

is the channel width

H

is the channel depth

The derivation of Equation 6.23 is presented in Chapter 4. Now, the horizontal velocity, vh, is related to the flow rate, Q, and channel geometry by:

vh =

3

Q BH

=

Cd B 2 g H 2 BH

1

Substituting for H 2 from Equation (6.23) yields: ©2000 Assoc.Prof R.J.Keller

1

= Cd 2 g H 2


(6.24)

Section 6


1

vh = Cd 2 g 

∴

vh( max) vh( min)

Q

3



(6.25)

Cd 2 gB 1

= 

6 -24

Qmax Qmin

3



(6.26)

Now, a typical value for the ratio of maximum to minimum flow rates is about 5. Substitution of this ratio into Equation (6.26) yields a corresponding value 1

for the ratio of maximum to minimum velocities of 5 3 = 1.71. If 0.2 m/sec is chosen for the value of vh(min), the corresponding value for vh(max) would be 0.342 m/sec, which would be unacceptably large. Accordingly, the shape of either the channel or the weir must be modified to maintain a satisfactory horizontal velocity. Modification of Channel Shape:

The issue to be resolved is whether or not it is possible to develop a channel shape such that the horizontal velocity remains constant for all flow rates. It is assumed that the channel discharges into a rectangular control section, such as a long-throated or Parshall flume. Such a device acts as a water level control and a flow measurement device. The analysis that follows is generally applicable to any rectangular crosssection. The analysis specifically makes use of the properties of a longthroated flume because it is widely used in practice and the analysis of the flume has been previously presented in Chapter 4. As shown by Equation (4.39), the flow through a long-throated flume may be expressed in the form: 3 2  2  Q =    g bc H1 2 3  3

where bc H 1

(6.27)

is the throat width is the upstream head

Differentiation of Equation (6.27) yields: dQ =

2 3

1

gbc H1 2 dH1

(6.28)

Now, within the channel, the horizontal velocity, vh, is given by: vh =


Q wH 1


(6.29)

Section 6


6 -25

or: Q = vh wH1

where w

(6.30)

is the channel width

Differentiation of Equation (6.30) yields the flow through an elemental horizontal strip of width w in the channel in the form: dQ = vh wdH1

(6.31)

Equating the right hand sides of Equations (6.28) and (6.31) yields: 2 3

1

gbc H1 2 dH1 = vh wdH1

(6.32)

Solution of Equation (6.32) for w yields: w=

2 3

g

bc vh

1

(6.33)

H 1 2

or, noting that vh is constant: 1

w = constant x H 1 2

(6.34)

Equation (6.34) describes a parabola, indicating that a parabolic shape for the channel cross-section will ensure a constant value of vh, regardless of flow rate. Design Aspects:

To reduce construction costs, the parabolic shape is normally approximated with a trapezoid. As a minimum, one channel and a bypass should be installed. When the number of channels is determined, the maximum, average, and minimum flows in an individual channel can be determined. The system should be designed such that, when one channel is out of service, its flow is diverted to the other channels. The resulting emergency flow for each channel is based on the maximum flow into the set of grit chambers with one out of service. The four flows, Qemerg., Qmax, Qave., and Qmin., are used to design the shape and length of the grit channel. Other practical aspects are associated with the turbulence which occurs in the inlet and outlet zones of the chamber. These zones are illustrated schematically in Figure 6.9. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -26

Turbulence occurs in the inlet zone as the flow is established. A similar phenomenon occurs in the outlet zone as the flow streamlines turn upwards. To allow for this disturbance, a 25 – 50 % increase in the calculated settling length is applied. Typical design criteria for a channel-modified horizontal grit chamber are presented in Table 6.4. A schematic of a typical channel-modified horizontal grit chamber is presented in Figure 6.10. Design Parameter

Typical Values

Comments

Water depth (m)

0.6 – 1.5

Dependent on channel area and flow rate

Length (m)

3 – 25

Function of channel depth and grit settling velocity

Extra outlet

for

inlet

and 25 – 50 %

Based length

on

theoretical

Detention time at peak 15 – 90 flow (seconds)

Function of velocity and channel length

Horizontal (m/sec.)

0.2 m/sec is Malaysian Standard

Table 6.4:

velocity 0.15 – 0.4

Typical Design Criteria for Channel-Modified Grit Chamber

The design procedure for a channel-modified grit chamber is illustrated in Example 6.3. Example 6.3

Design a horizontal/constant velocity grit chamber for a hydraulic load of 2,000 PE. Consider only the ADWF and the peak flow. Note: The water level within the chamber is controlled by a downstream long-throated flume which gives a depth of 205 mm at the peak design flow and 80 mm at ADWF. Maximum horizontal velocity is 0.2 m/sec Channel length > 18 x maximum water depth 3

Grit quantity is estimated as 0.03 m /ML of wastewater Grit collection channel to be cleaned out twice per week ©2000 Assoc.Prof R.J.Keller


Section 6


Figure 6.9:

Schematic of Settling Process in Grit Chamber

Solution

Average dry weather flow = 225 × 2,000 = 0.45 ML/day = 5.2l/sec Peak flow factor

= 4.7 × 2 −0.11 = 4.35

∴Peak flow

= 4.35 × 5.2 = 23 l/sec

Flow control gives depth of

205mm at peak flow 80mm at ADWF

(Consistent with long-throated flume of throat width 133mm) ©2000 Assoc.Prof R.J.Keller


6 -27

Section 6


6 -28

Calculate cross-sectional areas

ADWF:

Area =

0.0052 0.2

= 0026 . m2 Peak:

Area =

0.023 0.2

= 0115 . m2 Surface widths at each flow are now calculated Refer to Equations (6.27) and (6.33). 2

3

Q =   3

2

gwt y

& w=

2 3

3

2

g

(6.27)

wt

Transposing Eq. (6.33) wv h

wt =

2 3

gy

1

2

Substitute in Eq. (6.27) Q=

2 3

wyv h

∴Cross-sectional area =

2 3

wy

∴At average dry weather flow Surface width = A ×

=

3 2 y

0.026 × 3 2 × 0.08

= 0.49m



vh

y

1

2

(6.33)

Section 6


At Peak Flow Surface width =

×3 0115 . 2 × 0.205

=0.84m Length of chamber: > 18 × max. depth > 18 × 0.205 Use 3.7m Grit quantity: Based on average DWF Grit quantity = 0.45 × 0.03 3

= 0.014m /day

∴At twice weekly cleanout, grit accumulation = 0.014× ~ 4 = 0056 . m3 ∴Required cross-sectional area of grit collection channel =

0056 . 3.7

= 0015 . m2 Use grit collection channel 150mm wide × 110mm deep (gives some margin) Allow for freeboard (say, 200mm) Parabolic section to be approximated by trapezoid



6 -29

Section 6


6 -30

Figure 6.10: Schematic of Channel-modified Horizontal Constant Velocity Grit Chamber Modification of Downstream Control Weir:

For a rectangular grit chamber, the flow rate is given by: Q = vh By

where B y

(6.35)

is the chamber width is the flow depth in the chamber

The form of Equation (6.35) indicates that for vh to be constant, regardless of flow rate, the flow rate should be linearly proportional to the depth, y. This may be assured by using a downstream control weir characterised by a linear relationship between flow rate and head on the weir crest. Such a weir is the Sutro weir which is described and analysed in Chapter 4 of this Manual. For details and a worked example, refer to Section 4.4.4 and Example 4.5. 6.5.4

Vortex Grit Chamber

A schematic of a typical vortex grit chamber is shown in Figure 6.11. With reference to this figure, grit-laden flow enters the unit tangentially at the top. The resulting spiral flow pattern tends to lift the lighter organic particles while the mechanically induced vortex captures grit at the centre. The grit is then removed by air-lift or through a hopper. It should be noted that the grit



Section 6


6 -31

sump has a tendency to become compacted and clog. Sometimes provision is made for the use of high-pressure agitation water or air to clear the sump.

Figure 6.11: Schematic of Typical Vortex Grit Chambers (a) PISTA Unit (b) Teacup Unit

The adjustable rotating paddles maintain the proper circulation within the unit for all flows. However, attention should be paid to the tendency for these paddles to collect rags. Vortex grit chambers are highly energy-efficient. The head loss across the unit is minimal when operating correctly and unclogged. American practice indicates a value of 6 mm, although an allowance of 100 mm is recommended. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -32

Vortex grit chambers have the great advantage that they are very compact. Their design is usually proprietary so that manufacturers will usually produce a suitable unit to accommodate stated performance specifications. Manufacturers’ specifications will provide information on the maximum water depth within the chamber. 6.5.5

Aerated Grit Chamber

Aerated grit chambers are commonly used in medium to large sewage treatment plants. The introduction of air through a diffuser, located on one side of the tank, induces a spiral flow pattern in the sewage as it moves through the tank, as shown in Figure 6.12. Correct positioning of the tank inlet and outlet directs the flow perpendicular to the spiral roll pattern. Inlet and outlet baffles are normally installed to dissipate energy and minimise short-circuiting. Head loss across the chamber is minimal.

Figure 6.12: Helicoidal Flow Pattern in an Aerated Grit Chamber

The roll velocity is set so that it is sufficient to maintain lighter organic particles in suspension while allowing heavier grit particles to settle. Because conditions change with flow rate, the air supply is adjustable to provide the optimum roll velocity. A further advantage of the introduction of air is that the sewage is freshened, leading to a notable reduction in odour. If desired, the chamber can be used for ©2000 Assoc.Prof R.J.Keller


Section 6


6 -33

chemical addition, mixing, and/or flocculation ahead of primary treatment. Grease removal may be achieved with a skimmer. If correctly designed, an aerated grit chamber with a minimum hydraulic detention time of 3 minutes will capture about 95% of grit larger than 0.2 mm when operating at its peak flow. The usual range of design specifications is given in Table 6.5. The design of an aerated grit chamber is illustrated in Example 6.4. Example 6.4

Design an aerated grit chamber for a hydraulic load of 20,000 PE. Note: The minimum detention time at peak flow is 3 minutes The width to depth ratio is 2:1 The length to width ratio is 2:1 3

Grit quantity is estimated as 0.03 m /ML of wastewater The aeration requirement is 10 litres/sec/m length of tank

Design Parameter

Range of Values

Depth

2–5m

Length

8 – 20 m

Width

2.5 – 7 m

Width:Depth Ratio

1:1 – 5:1

Length:Width Ratio

3:1 – 5:1

Minimum Detention Time

2 – 5 minutes 3

Air Supply

0.25 – 0.75 m /min/m

Diffuser Distance from Bottom

0.6 – 1.0 m

Transverse Roll Velocity

0.6 – 0.75 m/sec

Table 6.5:

Comments

Varies widely

2:1 typical

3 minutes typical 3

0.45 m /min/m typical

Typical Design Specifications for an Aerated Grit Chamber



Section 6


Solution

Average DWF = 20,000 × 225l/day = 4.500m3/day =52l/sec -0.11

Peaking factor = 4.7 × 20 = 3.38

∴Peak flow

= 52 × 3.38 = 176l/sec

Grit chamber volume: Minimum detention time at peak flow = 3minutes = 3 × 60 = 180seconds

∴Required volume = 0.176 × 180 = 31.7 = 32m W D

L

= 2,

W

3

=2

∴Volume = D × W × L = 32 W = 2D,

L = 2W = 4D

∴D × 2D × 4D = 32 Dimensions

∴D = 1.6m ∴W = 3.2m L = 6.4m

Aeration requirement

10l/sec/m length = 10 × 6.4 = 64 l/sec Grit quantity

Based on average flow rate, 3 = 4.5ML/day × 0.03m /ML = 135l/day Note: Means should be provided to vary the air flow rate to control grit removal rate and grit cleanliness.



6 -34

Section 6

6.6


6 -35

Hydraulics of Clarifiers 6.6.1

Preliminary

Clarifiers are essentially sedimentation tanks and are used as a part of both primary treatment and secondary treatment processes. They may be rectangular, square, or circular in shape. A schematic of a typical circular clarifier has been presented in Figure 6.1. The flow enters at the centre of the tank and settlement takes place as the flow moves outwards and rises. The effluent is collected in a channel or launder, which then conveys the flow to an exit channel or pipe. This section emphasises the hydraulic aspects of the design of clarifiers. Design guidelines are first presented and the basic design procedure is reviewed. The important procedure for the design of the launder is then discussed. Finally, a design example is presented to aid understanding. 6.6.2

Design Guidelines

Design guidelines for primary and secondary clarifiers vary significantly from country to country. Typical guidelines from American practice are presented in Table 6.6. Parameter

Value Primary Clarifiers

Surface loading rate For average dry weather flow For peak flow conditions

32 - 49 m /m /day 3 2 49 - 122 m /m /day

Sidewater depth

2.1 – 5 m

Weir loading rate

125 – 500 m /m/day

3

2

3

Secondary Clarifiers

Surface loading rate For average dry weather flow For peak flow conditions

16 – 29 m /m /day 3 2 41 - 65 m /m /day

Sidewater depth

3.0 – 5.5 m

Floor slope

Nearly flat to 1:12

Maximum diameter

46 m

Table 6.6: ©2000 Assoc.Prof R.J.Keller

3

2

Typical Design Guidelines for Circular Clarifiers Manual of Practice Hydraulics of Sewerage Systems

Section 6


6 -36

Primary clarifiers are designed more conservatively if sedimentation is the only treatment and if activated sludge is being returned to the primary clarifier. Rectangular clarifiers are generally designed under the same criteria as circular clarifiers. Typical length to width ratios for rectangular primary clarifiers range from 3:1 to 5:1, although many existing tanks are characterised by ratios of between 1.5:1 and 15:1. A well designed and operated primary clarifier should be capable of removing between 50 and 65% of the influent total suspended solids. 6.6.3

General Design Principles

Clarifiers are designed to remove the maximum amount of settleable solids quickly and economically. The design objective is to provide sufficient time under quiescent conditions for maximum settling. If all solids were discrete particles of uniform size, density, and shape the removal efficiency of the tank would be dependent on the surface loading rate only as discussed in Section 6.3. It was also shown in Section 6.3 that the depth of the tank would have little influence on the removal efficiency provided horizontal velocities were maintained below the scouring velocity. However, the solids are not of a regular character and the conditions under which they are present range from total dispersion to complete flocculation. In practice, the bulk of the finely divided solids reaching primary sedimentation tanks are incompletely flocculated and are susceptible to further flocculation. Flocculation is aided by the eddying motion of the fluid within the clarifier. It proceeds through the coalescence of fine particles at a rate that is a function of their concentration and of their natural ability to coalesce upon collision. Thus, the longer the process continues, the more complete the coalescence becomes. For this reason, the detention time within the clarifier is a consideration in the design process. It should be noted, however, that the mechanics of flocculation are such that, as the time of sedimentation incless and less coalescence of the remaining particles occurs. Accordingly, from the point of view of settling, there is a practical limit on the effective detention time of the sewage. Primary and secondary clarifiers are normally designed to provide a detention time of between 1.5 and 2.5 hours, based on the average flow rate. It is noted that the design criteria for Malaysian systems incorporate a time of 2 hours based on the peak flow rate. Recalling the discussion of Section 6.3.4, the detention time is given by the equation: t =

H Q A

where Q/A is the surface loading rate. ©2000 Assoc.Prof R.J.Keller


(6.14)

Section 6


6 -37

Schematics of rectangular (horizontal flow), circular (radial flow), and square (upflow) clarifiers are presented in Figures 6.13, 6.14, and 6.15.

Figure 6.13: Schematic of Rectangular Sedimentation Tank

Figure 6.14: Schematic of Circular Clarifier

Rectangular tanks are commonly used for primary sedimentation. They occupy less space than circular tanks and can be economically built side by side with common walls. Circular tanks require careful design of the inlet stilling well to achieve a stable radial flow pattern without causing excessive turbulence in the vicinity of the central sludge hopper. Inlet design is considered in subsequent paragraphs. Upflow tanks typically have deep hopper bottoms and are common in small treatment plants. Their primary advantage is that sludge removal is carried out 0 entirely by gravity. The steeply sloping sides – typically 60 – concentrate the sludge at the bottom of the hopper. A significant disadvantage is that hydraulic overloading may cause major problems because any particles with a settling velocity less than the surface loading rate will not be removed, instead escaping with the effluent. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -38

Figure 6.15: Schematic of Upflow Clarifier

The primary design parameters are the surface loading rate and the detention time, both of which are normally specified in local design criteria. Following the specification of these parameters, the dimensioning of the tank then proceeds as follows: Tank Surface Area, A =

Q Q A

Tank length or Diameter, L or D = α A where α =

α =

L

4 4

π

(6.36)

(6.37)

for rectangular tanks

for circular tanks

The forward velocity is also an important aspect of the design of rectangular tanks. If this is excessive, scouring and re-suspension of the sludge will result. The forward velocity is given by: vh =

Q WH

(6.38)

Incorporating Equation (6.14) for the detention time, vh = ©2000 Assoc.Prof R.J.Keller

L t


(6.39)

Section 6


6 -39

It is evident from Equations (6.38) and (6.39) that the forward velocity influences the choice of length to width ratio. The maximum forward velocity to avoid the risk of scouring settled sludge is 10 to 15 mm/sec, indicating that the ratio of length to width should preferably be about 3:1. Values of L/W in practice range between 3 and 6. The Malaysian Draft Guidelines specify a value of 3. Design curves to aid in the determination of the tank geometry have been presented by Barnes (1981) and should be consulted for further information. Weir Loading Rate

The weir loading rate is defined as Q/Lw where Lw is the length of the outlet weir. If this value is too high, the approach current generated by the weir will extend upstream into the settling zone, creating a potential disruption of the 3 flow pattern. A weir loading rate of between 100 and 200 m /m/day is typically specified. Achieving this value is a particular problem for rectangular tanks which is usually overcome by utilising multiple suspended weir troughs. In circular tanks, the weir loading rate associated with a perimeter weir is normally satisfactory at high flows. At low flows, however, difficulties may arise from a weir loading rate which is too small because the consequent very small flow depths over the weir make the tank flow pattern very sensitive to errors in weir levelling. This problem may be overcome by constructing the perimeter weir as a saw-tooth weir – or multiple V-notch – to increase the flow depth. The issues of surface loading rate, detention time, and weir loading rate are illustrated by Examples 6.5 and 6.6. Example 6.5

Two primary clarifiers are 26 m in diameter with a 2.1 m side water depth. Single effluent weirs are located on the peripheries of the tanks. For a 3 wastewater flow of 26,000 m /day, calculate: a.)

The surface loading rate

b.)

The detention time

c.)

The weir loading rate

Solution

Surface area of each clarifier =

π D 2 4

=

π × 26 2 4

= 530m 2 ©2000 Assoc.Prof R.J.Keller


Section 6


6 -40

∴Total surface area = 530 × 2 = 1,060m

∴Total volume

= 1,060 × 2.1 = 2,230m

a.)

2

Surface loading rate

=

=

3

Q A

26,000 1060 ,

= 24.5m 3 / m 2 / day

b.)

Detention time =

=

Volume Flow rate 2,230 26,000

× 24

= 2.06 hours c.)

Weir loading rate

=

=

=

flow rate weir length 26,000 2 × π D 26,000 2 × π × 26

= 159m 3 / m / day Example 6.6

Determine the size of two identical circular final clarifiers for an activated 3 sludge system with a design flow of 20,000 m /day, and a peak hourly flow of 3 32,000 m /day. 3

2

Note: The maximum surface loading rate is 33 m /m /day at design flow and 66 m3/m2/day at peak flow. Minimum detention time at design flow is 2 hours 3

Maximum weir loading rate at design flow is 125 m /m/day



Section 6


Solution

At design flow, surface area required for each tank 20,000m 3 / day

=

3

2

2 × 33m / m / day

= 303m 2

Check peak overflow rate 32,000

=

2 × 303

= 53m 3 / m 2 / day < 66m 3 / m 2 / day

(OK)

Tank diameter

π D 2 4

= 303

∴ D = 

303 × 4

π

1



2

= 19.6 m Detention time =

∴

Tank volume Flow rate Area × Depth Flow rate

∴ Depth >

> 2 hours

2 × 10,000 303 × 24

> 2.75m Make depth 3.5m (Recommended for tank diameter > 15m, depth should be 3.4m). This will also give a reasonable detention time at peak flow rate. Weir loading rate

=

flow rate weir length

For single sided weir, weir loading rate/tank ©2000 Assoc.Prof R.J.Keller


6 -41

Section 6


=

6 -42

10,000

π × 19.6

= 162m 3 / m / day > 125m 3 / m / day

(No good)

∴Use an inboard weir channel with entry on both sides. Set weir channel at a diameter of 18m.

∴Weir loading rate =

10,000 2 × π × 18

= 88m 3 / m / day < 125m 3 / m / day

(OK)

Two tanks, Diameter

19.6m

Depth

3.5m

(20m?)

Inboard weir set on diameter of 18m. Tank Inlets

Sedimentation tank inlets must be designed to distribute the flow as uniformly as possible so that the best possible flow pattern is maintained. The influent jet has a high amount of kinetic energy that must be dissipated. For rectangular tanks, various baffled inlet arrangements have been used which are effective for energy dissipation and flow distribution. Typical arrangements are shown schematically in Figure 6.16. With circular tanks, the radial flow from the inlet is inherently less stable than the horizontal flow in a rectangular basin. Careful design is needed to achieve a stable radial flow pattern. Typical arrangements are shown in Figure 6.17 for (a) side feed, (b) vertical pipe feed, and (c) slotted vertical pipe feed. In all cases, the primary design principles are that energy must be dissipated and the flow distribution must be uniform.



Section 6


6 -43

Figure 6.16: Schematics of Typical Sedimentation Tank Inlets

Figure 6.17: Centre-feed Inlets for Circular Clarifiers: (a) Side Feed, (b) Vertical Pipe Feed, (c) Slotted Vertical Pipe Feed



Section 6


6 -44

Effluent Launder Design

Rising wastewater in a clarifier flows over a weir into a channel or launder which, in turn, conveys the collected effluent to the exit channel. Flow in the launder is classified as spatially varied because the flow rate increases with distance along the launder. This characteristic requires the use of the momentum equation for its analysis, rather than the energy equation. The basic flow condition is illustrated schematically in Figure 6.18 which shows the flow spilling over the multiple V-notch weir into the launder. A full momentum analysis, including the effects of friction, has been presented by Droste (1997). A simplified approach is usually adequate and is presented herein.

Figure 6.18: Definition Sketch for Flow in a Launder

The first issue is the size of V-notch weir required. The individual V-notches are typically set out with a centre to centre spacing of between 150 and 300 mm. With the number of V-notches consequently established, the flow through each can be determined from: Q perV − notch =

where N

Q

(6.40)

N

is the number of V-notches.

The maximum height, H , over the weir is then determined from the standard V-notch weir equation – refer to Chapter 4.4.3, Equation (4.24): Q perV − notch =

8 15

Cd 2 g tan

θ 2

H

5

2

(6.41) 0

The discharge coefficient, C d , is a function of the notch angle, θ . For θ = 90 , C d has a value of 0.58. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -45

The head over the weir, calculated from Equation (6.41), should be increased by a safety factor of 15%. The next stage in the hydraulic design is to determine the maximum depth in the launder. First, the critical depth at the discharge point of the launder is calculated from:

yc = 

( qL)

1

2

3



(6.42)

2  4b g 

where q =

Q L

and L is the length of the weir (circumference of the tank) is the width of the launder

b

The depth at the upstream end of the launder is then calculated from: 2

H =  y + 2 c

where x =

L

2

2q x

1

2

gb 2 yc

2



(6.43)

for a circular basin.

The depth, H , calculated from Equation (6.43) should be increased by a factor of safety of 50% to allow for friction loss, freeboard, and a free fall allowance. The derivation of Equations (6.42) and (6.43) has been presented by Droste (1997), in which further refinement is provided by including friction loss. This refinement would enable the full longitudinal profile in the launder to be calculated. It should be noted that such a refined design is rarely justified because design practice usually ensures that the launder is hydraulically over-designed. The design of a launder is illustrated by Example 6.7 Example 6.7

Design the overflow weirs and launders (collection channels.) for the clarifiers of Example 6.6. Note: The critical condition is when the peak flow occurs with one clarifier out of service. The launder must be able to cope with the corresponding flow. Solution

Weir design One clarifier must handle peak flow. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -46

∴Peak weir loading rate =

32,000m 3 / day 2 × π × 18m

where: 2 represents the inflow on both sides 18 represents the diameter

= 283m 3 / m / day Assume that weir comprises V-notches with spacing of 25cm centre to centre. (This may need adjusting)

∴Total number of V-notches = 2×

π × D 0.25

Take D as 18m, even though it will be less for the inner ring and more for the outer.

∴Total number of V-notches = 2×

π × 18 0.25

= 452

∴Flow per notch =

32,000 3, 600 × 24

×

1 142

= 0.00082m 3 / sec

0

Now, for each V-notch, notch angle is 90 and Cd =0.58. Q=

8 15

Cd 2 g tan

∴ H = 

θ 2

H

15 × 0.00082 8 × 0.58 19.6

5

2

2

5



= 0.051m



Section 6


A safety factor of 15% is normally appropriate.

∴Allow for water depth over notch of 1.15 × 0.051 = 0.059m = 60mm

∴Width of V-notch at the top = 60mm ×2 = 120mm. ∴Weir design as follows: Launder design

Q is discharge/unit length of launder = weir loading rate × 2 (because launder is fed from both sides)

∴q =

283 3, 600 × 24

× 2m 3 / m / sec

= 0.0066m 3 / m / sec Assume a launder width

Try 500mm Calculate depth at launder discharge point

y c = 

(qL)

1

2

3



2  4b g 

=

( 0.0066 × π × 18) 2 4 × 0.52 × 9.81

1

3



=0.243m



6 -47

Section 6


6 -48

Calculate maximum depth in launder at upstream end 2 H =  y +

(Note: x =

2q 2 x 2 gb 2 y

π D 2

0.5



)

2  π × 18  2  2 × 0.0066 ×    2 2  ∴ H =  0243 + . 2  9.81 × 0.5 × 0.243    

1

2

=0.419m Increase this depth by 50% to allow for friction loss in the launder, freeboard, and free-fall allowance.

∴Total depth to be provided in launder = 0.419 × 1.5 = 0.629, say 0.65m

∴Launder depth below vertex of V-notch weirs = 0.65m Launder width = 0.50m

6.7

Sludge Hydraulics 6.7.1 7.1

Prel relimi iminary

Sludge produced in sewage treatment plants must be conveyed from one plant point to another. The conditions of the sludge range from the consistency of water or scum to a thick sludge. It may also be necessary to pump sludge offsite for long distances for treatment and disposal. For each type of sludge and pumping application, a different different type of pump may be needed. The primary issues of concern are the type of pump to use, the computation of head loss in pipes carrying sludge, and other practical hydraulic aspects. These issues are examined in this section. Types of pumps are briefly discussed. Simplified computations, suitable for short lengths of pipe, are then presented. The application of rheology to head ©2000 Assoc.Prof R.J.Keller


Section 6


6 -49

loss computations for long-pipe calculations is then presented. Finally, practical aspects of sludge piping are briefly covered. It should be noted, in particular, that details of sludge processes are not covered because they are outside the scope of hydraulics concern. Specialist texts should be consulted for these details. 6.7. 6.7.2 2

Slud Sludge ge Pumpi umpin ng

Although specifying a single type of pump to handle all sludges within a treatment plant is an attractive idea, the wide range of conditions imposed on such service normally exceeds the capabilities of a single type of pump. Fortunately, many types of pump are available to the design engineer. Types of pumps most frequently used to convey sludge include the plunger, progressive cavity, centrifugal, torque flow, diaphragm, high-pressure piston, rotary lobe, and screw lift pumps. Specialist literature should be consulted for details on each. The application of different types of pump is summarised in Table 6.7. Commonly, centrifugal pumps of non-clog design are used. Problems arise, however, over choosing the most appropriate size. These problems occur because, at any given speed, centrifugal pumps operate well only if the pumping head is within a narrow range. Because of the variable nature of sludge, pumping heads may vary significantly. Selected centrifugal pumps must have sufficient clearance to pass the solids without clogging, but have a small enough capacity to avoid pumping a sludge diluted by large quantities of sewage overlying the sludge blanket. It is impractical to throttle the discharge to reduce the capacity because of frequent stoppages. For this reason, it is essential that these pumps be equipped with variable speed drives. Where the application involves high pressure, multiple pumps may be used and connected connected in series. Usually the consistency of untreated primary sludge changes during pumping with the most concentrated sludge being pumped first. Later, the pump must handle a dilute sludge which has essentially the same hydraulic characteristics as water. This change in characteristics causes a centrifugal pump to operate farther out on its characteristic curve. It is necessary that the pump motor is sized for the additional load and that a variable speed drive is used to reduce the flow under these conditions. It should be noted that if the pump motor is not sized for the maximum load when pumping water at top speed, it will go on overload or be damaged if overload devices devices do not function or are set too high. In determining the operating speeds and motor power required for a centrifugal pump handling sludge, it is important that system curves be determined for the densest sludge anticipated, the average conditions, and water. These system curves should be plotted on the pump characteristic crves for a range of available speeds. ©2000 Assoc.Prof R.J.Keller


Section 6


Principle

Kinetic (rotodynamic) pumps

6 -50

Common Types

Typical Applications

Nonclog mixed-flow pump,

Grit slurry, incinerator ash slurry Unthickened primary sludge

Recessed-impeller pump (vortex pump, torque-flow pump)

Return activated sludge

Screw centrifugal pump

Waste activated sludges from attached-growth biological processes

Grinder pump

Circulation of anaerobic digestor Drainage, filtrate, and centrate Dredges on sludge lagoons

Positivedisplacement pumps

Plunger pump

Waste activated sludge

Progressing cavity pump

Thickened sludges (all types)

Air-operated diaphragm pump

Unthickened primary sludge

Rotary lobe pump

Feed to dewatering mahines

Pneumatic ejector

Unthickened secondary sludges

Peristaltic pump

Dewatered cakes

Reciprocating piston Other

Air lift pump

Return activated sludge

Archimedes screw pump

Tabl Tablee 6.7: 6.7:

Slud Sludge ge Pum Pump p App Appli lica cati tion onss by by Pri Princ ncip iple le

The intersection of the pump curves with the system curves at the desired capacity yields the maximum and minimum speeds required for a particular pump. The intersection of the maximum speed pump curve with the system curve for water permits the determination of the power required. For the determination of hours of operation, average speed, and power costs, the intersection of the pump curve with he system curve for average conditions is appropriate. ©2000 Assoc.Prof R.J.Keller


Section 6

6.7.3


6 -51

Head Loss Determination

It is clear that the procedures in the previous section require an estimate of the head loss in the pumping lines. The head loss depends on the rheology (flow properties) of the sludge, the pipe diameter, and the flow velocity. It is known, further, that head osses increase with increased solids content, increased volatile content, and reduced temperatures. It is also known that, when the product of the percentage of volatile matter and the percentage of solids exceeds 600, difficulties in pumping sludge are often experienced. Dilute sludges such as unconcentrated activated and trickling filter sludges behave in a very similar manner to water. They are classified as “Newtonian” fluids. As such, the pressure drop is proportional to the velocity and the viscosity under laminar conditions, and to the square of the velocity under turbulent conditions. The head loss in pumping unconcentrated sludges may be between 10 and 25% greater than for water. Concentrated sludges, however, are non-Newtonian fluids. The pressure drop under laminar conditions is not proportional to the velocity and the viscosity is not a constant. Primary, digested, and concentrated sludges at low velocity are characterised by a plastic phenomenon whereby a definite pressure is required to overcome resistance and start the flow. The resistance then increases approximately with the velocity up to a velocity of about 1.1 m/sec, defining the upper limit of the laminar flow regime. Above about 1.4 m/sec, the flow may be considered to be turbulent. Within the turbulent range, the head losses for well-digested sludge may be two to three times greater than for comparable water velocities. For primary and concentrated sludges, the losses may be substantially greater. Two approaches for calculating head losses are considered in the following. A simplified approach is considered first, which is particularly suitable for short pipe lines. A more complex method is then discussed which uses the sludge rheology and is suited to head loss calculations in long pipe lines. Simplified Approach

The simplified approach is used to compute head losses in short pipe lines. The accuracy is adequate, especially for solids concentrations less than 3% by weight. Firstly, the head loss of water at the same flow rate is determined, using any one of the Darcy-Weisbach, Hazen-Williams, or Manning equations. This head loss is then multiplied by a factor, k, obtained from empirical curves for a given solids content and sludge type, or for a given velocit y and solids content. The first method is suggested when the pipe velocity is greater than 0.8 m/sec, thixotropic behaviour is not considered, and the pipe is not obstructed by grease or other material. Figure 6.19 presents the multiplication factor, k, as a function of solids concentration for digested sludge and for untreated primary and concentrated sludges respectively. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -52

Figure 6.19: Head Loss Multiplication Factor for Different Sludge Type and Concentration

The second method is less restrictive in its application and involves only the pipe velocity and solids concentration in determining the multiplication factor. Figure 6.20 presents the corresponding relationship for k. Application of Rheology to Head Loss Computations

Where sludge must be pumped over long distances, the accuracy of the estimates of head loss becomes more important because of their increased impact on the design of pumping needs. For this reason, the head loss computations should take account of the rheological properties of the sludge. In the discussion following, a method is described which uses similar concepts to the Darcy-Weisbach method, but with modifications to allow for sludge properties. Sludge behaves like a Bingham plastic – ie it exhibits a linear relationship between shear stress and flow only after flow begins. A Bingham plastic is described by two constants, the yield stress, s y, and the coefficient of rigidity, η . Typical ranges of values for these two constants are presented in Figures 6.21 and 6.22 respectively. It should be noted, however, that published data are highly variable. If considered important, pilot studies should be undertaken to determine the rheological data for specific applications.



Section 6


6 -53

Figure 6.20: Head Loss Multiplication Factor for Different Pipe Line Velocities and Concentrations



Section 6


6 -54

Figure 6.21: Range of Design Values for Yield Stress as a Function of Percentage Sludge Solids

Figure 6.22: Range of Design Values for Coefficient of Rigidity as a Function of Percentage Sludge Solids ©2000 Assoc.Prof R.J.Keller


Section 6


6 -55

Following the determination of the yield stress and the coefficient of rigidity, two dimensionless numbers are used to determine the pressure drop as follows: Reynolds Number Re = where ρ

ρ VD η

(6.44) 3

is the density of the sludge (kg/m )

V

is the average velocity in the pipe (m/sec)

D

is the pipe diameter (m)

Hedstrom Number He =

D2 s y ρ

η 2

(6.45)

The friction factor for the pipe-sludge system is then determined using the graph in Figure 6.23.

Figure 6.23: Friction Factor for Sludge Analysed as a Bingham Plastic

The pressure drop in the pipe is then calculate d from:

∆ p =

2 f ρ LV 2 D

(6.46)

Equation (6.46) can be readily shown to be a form of the Darcy-Weisbach equation. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -56

The equations and graphs presented above apply to the entire range of laminar and turbulent flows. It should be noted, however, that Figure 6.23 does not include any allowance for pipe roughness. To allow for pipe roughness, it is recommended that, in addition to the above procedure, the pressure drop should be calculated using a standard procedure for water. If this process gives a higher pressure drop than that given by Equation (6.46), roughness is dominant and the pressure drop given by the water formula will provide a reasonably accurate estimate of pressure loss. However, where use of the water formula is indicated, for worst case design conditions, a safety factor of 1.5 is recommended. The procedure for calculating head loss, including the sludge rheology, is illustrated in Example 6.8. Example 6.8

A pipeline of length 1,000 m and diameter 200 mm conveys untreated (raw) sludge at an average flow rate of 40 l/sec. Calculate the head loss in the pipeline. Analysis of the sludge indicated the following rheological data: 2

Yield stress:

sy = 1.1 N/m

Coefficient of rigidity:

η = 0.035 kg/m/sec

Specific gravity:

S.G. = 1.01

Solution

Pipe flow velocity Q V = A 0040 .

=

π ×

0.2 2

4 =1.27m/sec Sludge density = 1,000 × 1.01 3 = 1,010kg/m Reynolds Number ρ VD Re =

=

µ 1,010 × 127 . × 0.2

0035 . 3 =7.33 × 10 ©2000 Assoc.Prof R.J.Keller


Section 6


6 -57

Hedstrom Number 2 D s y ρ He = 2

µ

=

0.2 2 × 11 . × 1,010

2 0035 . = 3.63 × 10 4

From friction factor diagram f = 0.007

∴ ∆ p = =

2 f ρ LV 2

D 2 × 0.007 × 1,010 × 1,000 × 127 . 2 2

0.2

=114.03kN/m

∴Pressure loss in metres of water

=

∆ p 114.03 × 1,000 = γ 9.81 × 1,000

= 1162 . m

6.7.4

Sludge Piping

In sewage treatment plants, sludge piping should normall y not be less than 150 mm in diameter to prevent blockages. In exceptional circumstances, glasslined pipe of smaller diameter has been used successfully. Because of their greater risk of blockage, gravity sludge withdrawal lines should not be less than 200 mm in diameter. Pump connections should not be smaller than 100 mm in diameter. Instead of elbows in the line, it is good practice to install cleanouts in the form of plugged tees or crosses so that the lines can be rodded if necessary. Velocities in the piping should be between 1.5 and 1.8 m/sec and, not withstanding the minimum sizes, the pipe should be sized to maintain these velocities. Grease has a tendency to coat the inside of piping used for primary sludge and scum. Most often, this is much more of a problem in large sewage treatment plants than in small plants. Grease accumulation results in a decreased effective diameter and a consequent large increase in pumping head. The buildup of head occurs more slowly in systems where more dilute sludges are pumped. In some plants, specific provision is made for melting grease by circulating hot water, steam, or digester supernatant through the main sludge lines. Friction losses are usually relatively low in sewage treatment plants because the pipe lengths are relatively short. There is, accordingly, little difficulty in providing an ample safety factor. ©2000 Assoc.Prof R.J.Keller


Section 6


6 -58

In long sludge lines, however, special design features should be considered. The provision of two parallel pipes should be considered, unless a single pipe shut down for maintenance for several days will not create problems. Allowance for external corrosion and pipe loads should be considered. The provision of facilities to supply dilution water to the lines for flushing purposes may be necessary. Mechanical cleaning of the pipe lines may be necessary and provision should be made for the insertion of a pipe cleaner. Alternatively, or in addition, provision for steam injection for cleaning may be appropriate. Air relief and blowoff valves at high and low points, respectively, in the pipe line may be indicated and the likelihood of water hammer phenomena, consequent to pump and/or valve operation should be considered. A discussion of water hammer is presented in Chapter 5.

6.8

Effluent Disposal Hydraulics 6.8.1

Preliminary

After treatment, sewage is either re-used or disposed of into the environment. Disposal is by far the most common and, since this is a re-entry into the hydrological cycle, it can be seen as the first step in a very indirect and longterm re-use. The most common method of disposal is by discharge and dilution into ambient waters. It should be noted that another means of disposal is by discharge onto the land where the treated sewage seeps into the ground and recharges underlying groundwater aquifers. Part of this sewage also evaporates and, particularly in desert areas, the evaporated fraction can be substantial. Land application is not covered in this section. The single most important element of effluent disposal is the associated environmental impact. There is an associated regulatory framework which affects such issues as the selection of discharge locations, the selection of outfall structures, and the level of treatment required. Thus, sewage treatment and disposal are linked and cannot be considered in i solation. This section is designed to give an introduction to effluent disposal into natural waterways. The overall topic encompasses many areas including ater quality parameters, water quality standards and criteria, hydraulic transport processes such as advection and diffusion, and constituent transformation processes. No attempt is made in this section to grapple with the complex numerical mathematics associated with transport and transformation processes. Many excellent texts are available which cover these topics. In this section, the issues of river outfalls and ocean disposal are examined qualitatively and empirically and some simple design rules are introduced.



Section 6


6.8.1

6 -59

River Outfalls

Many existing effluent discharges into rivers are very poorly designed. Often they comprise open-ended pipes which achieve minimal initial mixing. In shallow streams, open ended discharges on the bank may fall directly onto the water surface, creating the potential for foaming problems. Such problems can often be eliminated by utilising a submerged discharge point, farther out into the stream. Where such rivers are navigable, however, outfall design requires special attention and is likely to be closely regulated. Rapid initial mixing of an effluent discharge into a river can be achieved with a multi-port diffuser. Such a structure discharges the effluent through a series of holes or ports along a pipe extending into the river. For shallow rivers, very rapid vertical mixing is achieved over the full river depth. Turbulent entrainment then draws river water into the effluent plume, promoting rapid dilution. This situation is shown schematically in Figure 6.24, which also shows a typical elevation of a riser.

Figure 6.24: Plan and Elevation Schematic of a Typical River Diffuser

The initial dilution, S , achieved in the near field, defined as being within approximately one diffuser length, is given by: S = ©2000 Assoc.Prof R.J.Keller

UHL

2 Q D

1 + 1 +

2Q DU D cosα U 2 LH




(6.47)

Section 6


where U

6 -60

is the river velocity

H

is the river depth

L

is the diffuser length

U D

is the discharge velocity through each port

α

is the orientation of the ports above the horizontal

The diffuser length, L, is often the most important parameter as it largely determines the cost of the structure. Equation (6.47) is used to determine the length of diffuser required to achieve a prescribed level of dilution. The equation is applicable to shore-attached as well as mid-river diffusers. The equation shows that high port discharge velocities increase dilution. However, care must be taken to ensure that there are no subsequent problems of scour or navigation. In practice, port velocities should not exceed 3 m/sec, although this guideline may be exceeded where circumstances warrant it and especially during infrequent high-flow events. Figure 6.25 shows a typical river diffuser arrangement. The port spacing adopted is typically of the same order as the water depth. At the outboard end of the diffuser, a large cleanout port is provided to facilitate flushing.

Figure 6.25: Schematic of a Typical Diffuser Outfall

The primary purpose of a multi-port diffuser is to distribute the flow evenly along the entire length of the structure. For this reason, the discharge per port should be as uniform as possible along the length of the diffuser. This is achieved by decreasing the diameter of the diffuser pipe in steps as shown in Figure 6.25. The detailed design to ensure an even flow distribution is based ©2000 Assoc.Prof R.J.Keller


Section 6


6 -61

on the so-called “manifold problem”, the details of which have been discussed by Fischer et al (1979). The use of Equation (6.47) in practice is illustrated with Example 6.9. Example 6.9

Determine the length and number of discharge ports for a multiport diffuser that will provide a near-field dilution of 10 when discharging a maximum flow 3 of 1.5 m /sec into a river. Under low flow conditions, the river water depth is 1.2 m and the current speed is 0.6 m/sec. Note: For the shallow water conditions prevalent under low river flow conditions, the maximum discharge velocity, U D, should be lower than the value of 3 m/sec, recommended in the notes, to reduce the risk of bottom erosion and hazards to boaters. A value of 2 m/sec is suggested. Because of the shallow depth, the ports will discharge horizontally in the same direction as the river flow. Solution

Calculate required diffuser length S =

UHL

2Q D

∴10 =

1 + 1 +

0.6 × 12 . × L 2 × 15 .

2Q DU D cos α 2

U LH

1 + 1 +



2 × 15 . × 2 ×1 0.6 2 × L × 12 .

Solve by trial L=18m Determine required number of ports Port spacing ≈ water depth 18 ∴ No. of ports = +1 12 . = 16 Determine port diameter Q = ( On. of ports ) ×

∴ D0 = 

4 × 15 .

π × 16 × 2

π D02 4 1



× U D

2

= 0.244m Nearest standard size = 0.25m Check required number of ports





Section 6


6 -62

Q

N =

2

D0

U × π ×

4

15 . ×4

=

2 × π × 0.25 2 =15.3 Select 15 ports

∴Port velocity Q

=

N × π ×

D

2

4 15 . ×4

=

15 × π × 0.25 2 = 2.04m/sec

(OK)

Dilution rate S =

0.6 × 12 . × 18  2 × 15 . × 2.04 × 1  1 + 1 +  2 × 15 . 0.6 2 × 18 × 12 . 

= 10.1

(OK)

∴ Diffuser length:

18m Number of ports: 15 Port diameter: 250mm Port spacing: 1.29m Port velocity: 2.04m/sec Dilution (near field): 100.1

6.8.3

Ocean Disposal

Oceans and large lakes are used for effluent disposal by many communities. Provided that the outfall structure is appropriately designed, such water bodies provide extensive assimilation capacity. Sewage effluent is typically carried to an offshore discharge point by a pipe or tunnel. The actual discharge may be through a single port or multi-port diffuser. The characteristics of the effluent plume are complicated by the density difference that exists between the lighter effluent and the denser sea water. The density of the effluent is dependent on its temperature and, to a lesser extent, on the suspended solids concentration. The configuration of a typical effluent plume in the ocean is shown in Figure 6.26.



Hydraulics of Sewage Treatment Plants Sec-6

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